Cofferdams: Uses, Types, Design, and Construction

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November 7, 2022

Table of Contents

A cofferdam is a structure used to keep soil and/or water out of an area where construction work must be done to a depth below the surface. The effects of water penetration must always be taken into consideration in any calculations, even though the total exclusion of water is frequently unnecessary and occasionally not practicable.

The cofferdam should always be taken into consideration by the designer when constructing a basement. The use of steel sheet piles as the main permanent structural wall can result in significant time and cost savings. The wall can be built to support vertical loads and water tightness can be achieved by using the right sealing technique.

The use of top-down construction should be taken into consideration when controlling ground movement is a particular concern. This will prevent secondary movement from happening when the lateral soil loading is transferred from the temporary supports, as they are removed, to the permanent structure. It will also ensure that movement of the top of the wall is restricted with the introduction of support at ground level prior to the commencement of excavation.

There are two main methods for designing cofferdams. Although single-skin structures are more frequently employed, double-wall or cellular gravity structures may be used for very large or deep excavations and marine works.

Requirements for a Cofferdam

The cofferdam’s design must adhere to the following requirements:

The structure must be capable of withstanding all of the different loads applied on it.
Pumping must be capable of controlling the amount of water entering the cofferdam.
The formation level must be stable and not be prone to uncontrollable heaving, boiling, or piping at any stage of construction.
The permanent structure or any other existing structure next to the cofferdam shall not be impacted by the deflection of the cofferdam walls and bracing.
It must be demonstrated that overall stability exists in the face of out-of-balance earth forces brought on by sloping terrain or probable slip failure planes.
The cofferdam needs to be the right size for the construction work that will be done inside of it.
The construction of temporary cofferdams must be done in a way that allows for the greatest possible quantity of construction waste to be recycled.

Planning a Cofferdam

Before starting the design of a cofferdam, the designer must have a clear understanding of their goals. In order for the design to account for all the load situations related to the construction and takedown of the cofferdam, the sequence of construction operations must be established. The designer can find the key design instances from this sequence and then compute the minimal penetrations, bending moments, and shear forces necessary to establish the pile section and length.

The designer should conduct a risk assessment of the effects of any departure from the intended sequence as part of the analysis of the construction activities. These deviations could take the form of excessive excavation at any stage, a failure to reach the required piling penetration, the installation of the support at the incorrect level, or the imposition of a significant surcharge loading from construction equipment or materials.

The site management should be informed and contingency plans should be prepared to minimize any risk if any stage of the cofferdam construction is particularly fragile. This will reduce the likelihood that critical conditions will materialize.

The bulk of cofferdams are built as temporary works, hence it might not be cost-effective to plan for every conceivable loading scenario. When evaluating the design instances, decisions will need to be made, typically including the site management. An example of this may be when evaluating the hydraulic loading on a cofferdam. Flood conditions are typically seasonal, therefore creating a cofferdam that will always keep water out may need a significant increase in pile size, strength, and construction.

The design philosophy may call for evacuating the cofferdam in a severe flood situation and letting it overtop and flood instead. Under these circumstances, the designer must account for overtopping, taking into account the impact of any trapped water on the stability after the flood subsides as well as the effect of the abrupt ingress of water on the base of the cofferdam.

The site area should be cleared before construction begins to make room for the installation of plant and guide frames. Excavation shouldn’t start until all the necessary machinery and supplies, including pumping equipment when needed, are on hand to support the piles.

The cofferdam and support frames should be checked when excavation is finished to make sure they are functioning as planned and to give as early a warning as possible of any safety-critical issues. In the UK, it is required by law to keep a written record of such monitoring, which is good practice.

Below are a few potential causes of failure, and it will be clear that many of them have to do with issues that could develop after the cofferdam is complete.

Common Causes of Cofferdam Failures

Cofferdam failure may have a variety of causes, but in reality, it is typically caused by one or more of the following:

The structure’s installation and design lacked attention to detail.
Failing to consider the range of probable water levels and conditions.
Failing to compare facts learned during excavation with design calculations.
Excessive excavation during any phase of the construction process.
The amount and quality of the framework used to support the loads is insufficient.
Loads on frame members that were not considered during the design, such as those caused by walkways, materials, pumps, and other structures supported by walings and struts.
Accidental damage to structural components that are left unrepaired.
Inadequate penetration to prevent heaving or pipework. Failure to take into account how pipes or heave will affect soil pressures.
A breakdown in communication between the designers of permanent and temporary works, as well as between designers and site management or site management and workers.

In many circumstances, the occurrence of several of the aforementioned conditions at once might result in failure, even though any one of them might not have been enough to produce the failure on its own.

Design of Cofferdams

The ensuing remarks are especially pertinent to the design and construction of cofferdams. In order to choose the proper geotechnical parameters for the soils in which the cofferdam is to be built, the longevity of the cofferdam construction must be evaluated. Since the cofferdam is a temporary structure, total stress values can often be employed.

However, the designer must determine each clay’s susceptibility to the quick attainment of a drained condition and, if in doubt, should examine the final construction using effective stress parameters. Generally speaking, it is advised to use effective stress strength parameters when designing cofferdams that will be in use for three months or longer. It may be suitable to utilize effective stress criteria for considerably shorter periods of time since the existence of silt laminations or layers inside clays can quickly lead to the attainment of drained conditions.

The types of cofferdams to be discussed in this section are;

single-skin cofferdams
double skin/walled cofferdams
Cellular cofferdams

Single skin Cofferdams

Sheet piles that are either internally propped or externally anchored are generally used to create single skin cofferdams. Depending on the kind of soil, the piles will need to be driven to a level that will create enough passive resistance even though they are just intended to provide support between frames and below the lowest frame. If there are at least two frames, the wall will still be stable if the cut-off of the piles below the excavation is insufficient to give the required passive support, and the pile beneath the lowest frame can be thought of as a cantilever.

Nevertheless, this will result in significant stresses in the lowest frame and is to be avoided whenever possible. In every situation, the penetration below formation level must be sufficient to prevent water from entering the excavation.

driving of sheet piles — Driving of sheet pile walls

Records should be kept when driving to look for any signs of the piles declutching. To prevent seepage in such a situation, grouting may be required behind the piles. The same restrictions that apply to cantilever retaining walls apply to cantilever pile cofferdams as well, particularly in terms of the retained height that can be achieved. When the cofferdam has very wide plan dimensions but a modest depth, slanted struts or external anchorages are frequently more cost-effective to include. However, it should not be overlooked that space beyond the cofferdam area is needed for the installation of external anchorages, and wayleaves may be needed to place the anchors beneath neighbouring homes.

A system of internal frameworks in the form of steel sections or specialized bracing equipment is often used for a cofferdam with a depth of more than 3 meters. In order to accurately represent the construction process, the design should be carried out in stages. Excavation and dewatering to just below the top frame level would typically be followed by the installation of the first frame. This process would be repeated for each additional frame.

It should be noted that in the case of cofferdams submerged in water, the forces experienced during dewatering and frame installation may be significantly greater than those present in the finished cofferdam. It is recommended to use a specific interlock sealant for cofferdams in water. It is frequently more efficient to install all of the framing underwaters before beginning the dewatering process when a cofferdam is to be utilized exclusively for the purpose of excluding water and the depth of soil to be excavated is merely nominal.

The ideal frame spacing for this construction style is shown below. The spacing causes the second and succeeding frames to be loaded almost equally.

Walls/double skin Cofferdams

Double wall cofferdams are made up of two parallel lines of sheet piles that are joined at one or more levels by a network of steel tie rods and walings. Sand, gravel, or crushed rock are common granular materials used to fill the area between the walls. The outer line of piles serves as the anchorage, and the exposed or inner wall is intended to serve as an anchored retaining wall. For this type of construction, U or Z profile sheet piles are ideal.

The wall should be analyzed as a whole as a gravity structure, and it will typically be observed that the width should not be less than 0.8 of the retained height of water or soil in order to ensure acceptable factors of safety against overturning and sliding. It is advised that the logarithmic spiral approach developed by Jelinek be used to assess the structure’s overall stability.

Transverse bulkheads should be installed to create strong points at the ends and in the middle of the structure to facilitate construction and contain any potential damage. A square or rectangular cell tied in both directions may make up the strong points. Both within and outside of the construction, the water regime is crucial. Weepholes should be installed on the inner side of the building close to the bottom of the exposed piles to allow for unrestricted drainage of the fill material, lowering pressures on the inner wall and preventing the fill’s shear strength from deteriorating over time.

Weepholes are only useful for small constructions, and it’s not always practicable to drain the fill completely. If necessary, wellpoints and pumps give a speedy alternative for drainage. However, any water pressure acting on the piles should always be taken into consideration by the designer. It is crucial to avoid using clay or silt as fill material, and any of this kind that is found within the cofferdam and is above the primary foundation stratum must be removed before the fill is added.

The piles must be driven deep enough into the ground to produce the necessary passive resistance, which is below excavation or dredging level. Under these circumstances, the structure will lean toward the side that was excavated, and the lateral earth forces on the side that was retained may be regarded as active. The penetration of the piling must also be sufficient to control the effects of seepage when cohesionless soils are present at or below the excavation level. The structure’s weight and any additional loading should be compared to the founding stratum’s bearing capacity.

This type of cofferdam is inappropriate since there is rock at the excavation level unless:

• The type of the rock makes it possible to drive sheet piles into it deeply enough.
• Tie rods can be mounted at a low height (probably underwater).
• The piles can be positioned and grouted into a prepared trench in the rock.
• Dowels inserted into rock sockets can be used to pin the pile toes.

Base friction and gravity forces should be enough to prevent overturning and sliding if the piles are driven onto hard rock or to a nominal depth below the dredged level. Depending on the degree of deflection, the lateral earth pressure on the retained side will be somewhere between at rest and active in this situation.

Because of the uneven distribution of vertical stresses within the cofferdam (owing to moment effects), the internal soil pressures operating on the outside walls are likely to be greater than active, hence the design should be based on pressures that are 1.25 times the active values.

Cellular Cofferdams

Self-supporting gravity structures known as cellular cofferdams can be formed into a variety of designs utilizing straight web sheet piles. After being forced together, the piles form closed cells that are filled with cohesionless material. The circular cells are joined together using manufactured junction piles and brief arcs to achieve wall continuity.

They only need a small amount of penetration to be stable, provided the foundation material is solid. Pile penetration will help in the vulnerable period before the fill has been deposited and the cell has become intrinsically stable by assisting in the resistance of any lateral stresses that may arise during construction.

Cofferdams made of cells are used to hold back large amounts of water or later-placed fill. They are frequently used to create quay walls, breakwaters, and dock closure cofferdams. The straight web pile section, in particular the interlocks, has been created to resist the circumferential tension that forms in the cells as a result of the contained fill’s radial pressure while also allowing for enough angular deflection to allow for the formation of cells with a useful diameter.

The steel can be placed in cellular construction such that the maximum tensile resistance is created across the profile because no bending moments are developed in the sheet piles. Because of this, the sections have very limited bending strength and cannot be used to build standard straight sheet pile walls. Tie rods and walings are not necessary.

Dispersive Soils

By

Ubani Obinna

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November 5, 2022

When the repulsive forces between clay particles in soils are greater than the attractive forces, deflocculation results, causing the particles to resist one another and form colloidal suspensions in the presence of relatively pure water. Such soils are known as dispersive soils. In terms of erosion control, there is a specific threshold velocity below which running water does not erode non-dispersive soil. Only water flowing with a specific amount of erosive energy can separate the individual particles, which otherwise stick to one another.

Contrarily, dispersive soil does not have a threshold velocity because the colloidal clay particles remain suspended even in still water, making these soils extremely prone to erosion and piping. Except for the possibility that soils with less than 10% clay particles may not have enough colloids to sustain dispersive pipes, there are no notable differences between the clay fractions of dispersive and non-dispersive soils. Dispersive soils have a moderate to high clay material concentration.

In comparison to regular soils, dispersive soils have a higher concentration of dissolved salt (up to 12%) in their pore water. Clay particles occur as aggregates and coatings around silt and sand particles in soils with high salt levels, and the soil is flocculated as a result.

When free salts are present in the pore water, the sodium adsorption ratio (SAR) is used to assess the contribution of sodium. Sodium adsorption ratio (SAR) reflects the proportion of sodium ions to the total amount of calcium and magnesium ions in water and is defined as:

SAR = [Na⁺] / [(Ca²⁺ + Mg²⁺)/2]^0.5

with units expressed in meq/litre of the saturated extract.

The exchangeable ions in the layers of adsorbed clay particles are correlated with the electrolyte concentration of the pore water. The type of clay minerals present may have an impact on this relationship in addition to the pH value. As a result, it is not always constant. An SAR value greater than 10 was regarded indicative of dispersive soils by Gerber and Harmse (1987), between 6 and 10 was considered moderate, and less than 6 was considered non-dispersive.

Sodium Adsorption Ratio (SAR)	Degree of dispersivenenss
< 6	Non-dispersive
6-10	Moderate
> 10	Dispersive soil

The key chemical determinant of dispersive behaviour in soils is the availability of exchangeable sodium. In terms of the exchangeable sodium percentage (ESP), this is stated as follows:

ESP = Exchangeable [(Na)/(Ca + Mg + K + Na)] x 100

where the units are expressed as meq/100 g of dried clay.

Elges (1985) suggested a 10% ESP cutoff point over which soils that have their free salts leached by seepage of reasonably pure water are susceptible to dispersion. ESP values exceeding 15% indicate very dispersive soils (Bell and Maud, 1994). At ESP values of 6% or below, it was discovered that those with low cation exchange values (15 meq/100 g of clay) were absolutely non-dispersive. Similar to this, soils with high cation exchange capacity numbers and a plasticity index higher than 35% swell to the point where dispersion is negligible.

Exchangeable Sodium Percentage (ESP)	Classification
< 6	Non-sodic
6-10	Sodic
10-15	Moderately sodic
15-25	Strongly sodic
> 25	Very strongly sodic

Regrettably, standard soil mechanics testing cannot distinguish between dispersive and non-dispersive soils. No single test can be relied on totally to detect dispersive soils, despite the fact that several specialized tests have been employed to identify them (Bell and Maud, 1994).

Physical and chemical testing can be used to categorize them. The former includes the pinhole test, the modified hydrometer or turbidity ratio test, the dispersion or double hydrometer test, and the crumb test. Craft and Acciardi (1984) discovered that the pinhole and crumb tests occasionally produced contradictory findings from the same soil samples.

Then, Gerber and Harmse (1987) demonstrated that when free salts were present in solution in the pore water, as is frequently the case with sodium-saturated soils, the crumb test, the double hydrometer test, and the pinhole test were unable to identify dispersive soils.

Dispersive Soils in Construction

Dispersive soils have been employed in the construction of earth dams, resulting in serious piping damage to embankments (Bell and Maud, 1994). After a rainstorm, deep gullies caused by severe erosion damage can develop on earthen embankments. Small leaks of muddy-colored water from an earth dam after the reservoir has been initially filled are signs of pipework.

There is a risk that the dam will fall as a result of the rapidly expanding pipes. Dispersive erosion can be brought on by initial seepage through an earth dam in regions with higher soil permeability, particularly in places where compaction may not be as effective, such as near conduits, up against concrete structures, and at the foundation interface; or by desiccation cracks, differential settlement cracks, or cracks brought on by hydraulic fracturing.

There is often no other viable option economically in many locations where dispersive soils are present but to employ these soils to build earth dams. Experience suggests that an earth dam should be sufficiently safe even if it is constructed with dispersive soils provided there is thorough construction control and includes filters.

Sodicity and Dispersive Soils

According to Vacher et al. (2004), dispersive soils, which typically contain more than 6.0% exchangeable sodium, are where tunnel erosion mostly occurs (ESP). In the past, these soils may have been referred to as solodic, solonetz, or solodized – solonetz (Doyle and Habraken, 1993). These soils are known as sodic soils or Sodosols (Isbell, 2002). Sodic or dispersive soil layers may also be present in other soil types, including Vertosols, Kurosols, and Kandosols.

Dispersion is the process by which individual clay platelets split from the aggregate when sodic soil comes into contact with non-saline water. Water molecules are pulled in between the clay platelets, causing the clay to inflate to such an extent that they separate from the aggregates.

Small aggregates appear to “dissolve” into a milky ring or halo when they are placed in a dish of distilled water. The clay platelets that were ejected from the clay aggregate are what make up this milky ring. Dispersed platelets are frequently so tiny that they are perpetually suspended, which explains why dams made of dispersive clays never settle and consistently have a “muddy” or “milky” appearance.

Even while sodic soils tend to scatter, it’s crucial to recognize that not all sodic soils do, and not all dispersive soils are sodic (Sumner, 1993). While organic matter, clay mineralogy, acidity, and high iron content may inhibit sodic soils (ESP > 6%) from dispersing, factors like silt and high magnesium content may encourage non-sodic soils (ESP 6%) to do so (Raine and Loch, 2003; Rengasamy, 2002).

Despite possessing less than 6.0% ESP, degraded kurosols in southern Tasmania are known to be dispersive. In addition, until the salt is leached from the soil profile, usually after subsurface drainage, saline soils that are also sodic do not scatter or behave like sodic soils (Rengasamy and Olsson, 1991).

Sodic soils swell but typically don’t disperse in water that is moderately electrolyte (salty) concentrated or somewhat saline. Clay platelets are not broken. Salts in the soil water lower the osmotic gradient between the clay platelets’ outside and inside, preventing the last stage of swelling that would otherwise lead to dispersion (Nelson, 2000). One of the most crucial defenses against gully erosion and soil erosion that sodic soils have is the maintenance of salts in the soil water.

References

[1] Bell F.G. and Maud R.R. (1994): Dispersive soils: A review from South African Perspective. Quarterly Journal of Engineering Geology and Hydrogeology 27:195-210
[2] Craft D., and Acciardi R. G. (1984): Failure of Pore-Water Analyses for Dispersion, Journal, Geotechnical Engineering Division, ASCE, Vol. 110, No. 4, Apr. 1984.
[3] Doyle R. and Habraken F.M. (1993): The distribution of Sodic Soils in Tasmania. Australian Journal of Soil Research 31 (6), 931-947.
Elges, H. F. W. K. (1985) ‘Dispersive soils: problem soils in South Africa-state of the art’, Civil Engineer in South Africa, 27(7):347-349.
[4] Gerber, F.A. and Harmse, H.J. von M. (1987): Proposed procedure for identification of dispersive soils by chemical testing. The Civil Engineer in South Africa, 29:397-399.
[5] Isbell R.F. (2002): ‘The Australian Soil Classification’. Australian Soil and Land Survey Handbooks Series Volume 4, (CSIRO Publishing: Collingwood, Vic.).
[6] Nelson P.N. (2000): Diagnosis and Management of Sodic Soils Under Sugarcane, BSES Publications.
[7] Raine S. R. and Loch R.J. (2003): What is a sodic soil? Identification and management options for construction sites and disturbed lands. In ‘Road, Structures and Soils in South East Queensland 29-30th’ (Department of Main Roads, Queensland).
[8] Rengasamy P. (2002): Clay dispersion, In Soil Physical Measurement and Interpretation for Land Evaluation, Australian Soil and Land Survey Handbook Series, Vol. 5, Eds McKenzie N., Couglan K, and Creswell H. pp 200-210. CSIRO publishing, Collingwood, Victoria.
[9] Rengasamy P. and Olsson K.A. (1991): Sodicity and soil structure. Australian Journal of Soil Research 29:935-952
[10] Sumner M.E. (1993): Sodic Soils: New perspectives. Australian Journal Soil Research 31:683 – 750.
[11] Vacher C.A., Loch R.J. and Raine S.R. (2004): Identification and Management of Dispersive Mine Spoils. Final Report for Australian Centre for Mining Environmental Research, Kenmore Queensland.

Ground Improvement Using Stone Columns

By

Ubani Obinna

-

November 4, 2022

Table of Contents

Stone columns are vertically positioned piles of compacted, gravel-sized stone particles used to enhance the performance of soft or loose soils. The stone can be compacted using impact techniques, such as vibroflots, impact compactors, falling weights, and so on. The technique is used to increase bearing capacity (up to 5 to 10 ksf or 240 to 480 kPa), decrease foundation settlements, improve slope stability, reduce seismic subsidence, reduce lateral spreading and liquefaction potential, allow construction on loose/soft fills, and prevent sinkholes from pre-collapsing in karst regions.

A stone column’s performance is greatly influenced by its diameter. Greater strengths are more likely in larger columns, which is linked to a higher area replacement ratio. The in-situ soft soil is typically partially replaced, resulting in a denser ground that is further improved by adding more fill material.

The moisture content of the nearby base soil also has an impact on the connection between the imported material and the in-situ soil. Since the confinement pressure of the surrounding soil is what generally causes stone columns to function, it is obvious that reduced confinement forces (resulting from wetter soils) will have an effect on the column strength.

Applicable Soil Types

The performance of soils is enhanced by stone columns in two ways: first, by densifying the surrounding granular soil, and second, by reinforcing the soil with a stiffer, higher shear strength column. Table 1 displays the anticipated improvement for various soil types. In most cases, it doesn’t matter how deep the groundwater is.

Soil Description	Densification	Reinforcement
Gravel and sand <10% silt, no clay	Excellent	Very good
Sand with between 10 and 20% silt and <2% clay	Very good	Very good
Sand with >20% silt and nonplastic silt	Marginal (with large displacements)	Excellent
Clays	NA	Excellent

Table 1: Expected Densification and Reinforcement Achieved with Stone Columns

In general, stone columns can be utilized to enhance the ground when lightweight constructions are built on poor soils. The installation of these columns might be done with the hard stratum in contact or floating above it. Extending it to the hard stratum is preferred. By using this installation method, there is no chance of columns punching through the soil. As a result, differential settlement is greatly reduced. A few test investigations of stone columns that were extended to the hard stratum revealed that the tested columns only bulged in the upper third of the column, thus demonstrating their appropriate load-bearing capability.

Rammed stone columns may be used to support light weights in the following situations, depending on the common local applications requiring improvement of the ground conditions for construction purposes: strip footings, houses (up to two stories), embankment support, storage tanks like oil tanks, and slope stability. Due to the high permeability of crushed aggregates, which promotes faster drainage, the technique would be of utmost importance in soft soils that are heavily saturated.

The procedure of Stone Column Installation

Stone column construction begins at the bottom of the treatment depth and is worked up to the surface. With the help of its weight, vibration, and often water jets in its tip, the vibrator penetrates the ground using the wet top feed technique. The firm soils may also be predrilled through if poor penetration is experienced.

The stone is placed around the vibroflot by a front end loader at the surface of the ground, and it falls to the tip of the vibroflot through the flushing water. The stone then falls around the vibroflot to the tip, filling the void created when the vibroflot is lifted, and the vibrator is then raised a few feet. The stone is subsequently compacted and displaced in lifts of 2 to 3 feet (0.75 to 0.9 m) when the vibroflot is periodically raised and lowered as it is extracted.

Stone column installation — Schematics of Stone column construction

Typically, the flushing water is sent to a settlement pond where it is permitted for the soil fines suspended in the water to settle. When using the dry bottom feed process, the vibroflot only needs its weight and vibrations to help it enter the earth. Predrilling may once more be performed if required or preferred. The subsequent steps are then similar, with the exception that the stone is sent through the tremie pipe to the tip of the vibroflot. It has been possible to treat at depths of up to 100 feet (30 meters).

In general, two mechanical techniques—vibration and ramming—are used to install these columns. Rammed columns are placed by first generating a pre-bored hole which is afterwards filled with a compacted material in many layers as opposed to vibrated columns which employ a vibratory probe to generate an opening for granular fill placement by either the displacement or the replacement method.

The difficulty of the installation is what distinguishes the two methods. Vibrated columns are more expensive than rammed columns because they require more advanced equipment and experienced labour. Figure 1 depicts how a typical rammed stone column is installed.

Equipment

The setup and equipment utilized when jetting water is used to advance the vibroflot are similar to VC. For a given project, the dry bottom feed method can be employed if water jetting is not desired. Stone is fed to the vibroflot’s tip through a tremie pipe that is affixed to the side of the vibroflot. A front-end loader fills a stone skip with stone on the ground, and a different cable lowers the skip to a chamber at the top of the tremie pipe.

Vibro piers are an application in particular. Short, tightly spaced stone columns are used in the procedure to form a strong block that will boost bearing capacity and lower settlement to acceptable levels. In cohesive soils where a predrill hole may be dug to its maximum depth and remain open, vibro piers are often built. The stone is rammed and vibroflotted into 1 to 2 foot (0.4 to 0.8 m) lifts before being compacted.

Materials

Although natural gravels and pebbles have been utilized, the stone is often a hard rock that has been graded and crushed. The modulus and shear strength of the column increase with the stone’s friction angle.

Design of Stone Columns

There are numerous analysis techniques available. One technique for static analysis involves generating weighted averages of the soil and stone column parameters (cohesion, friction angle, etc.). The weighted averages are then utilized in conventional geotechnical analysis techniques (bearing capacity, settlement, etc.). The treatment limits and foundation limits are often equivalent in static applications.

Stone columns have several advantages for liquefaction analysis, including densification of nearby granular soils, a decrease in cyclic stress in the soil due to the presence of stronger stone columns, and drainage of the excess pore pressure. In liquefaction applications, the treatment typically extends laterally outside the sections to be protected and covers the whole footprint of the structure, which is equal to two-thirds of the thickness of the liquefiable zone. This is required to prevent the treated region beneath the foundation from being negatively impacted by the nearby untreated soils.

Quality assurance and quality control

Location, depth, ammeter increments, and the quantity of stone backfill utilized are crucial construction factors to track and record. To gauge the improvement made in granular soils, post-treatment penetration tests can be done. With test footings as large as 10 feet square (3.1 meters) and loaded to 150% of the design load, full-scale load testing are increasingly prevalent.

Design of Masonry Walls | EN 1996-1

By

Ubani Obinna

-

October 30, 2022

Table of Contents

Durability and strength are the two main types of limit states relevant to the design of masonry structures and masonry walls. The selection of masonry units and mortars for certain construction types and exposure classes is a key component of design for durability. The compressive strength of the masonry f_d, is determined by;

f_d = f_k/γ_M

where;
f_k is the characteristic compressive strength of masonry
γ_M is the partial factor for materials.

The following product/material characteristics affect the characteristic compressive strength of masonry:

• Group number of masonry unit
• Normalised mean compressive strength of masonry unit
• Compressive strength of the mortar.

The partial factor for materials is a function of the following aspects:

• Category of (masonry unit) manufacturing control
• Class of execution control.

In Europe, a wide range of masonry units are used. Numerous characteristics of the units vary, including the percentage, size, and orientation of voids or perforations, as well as the thickness of webs and shells. Therefore, the EN 1996-1-1:2004 (EC 6) drafting panel had to come up with strategies that would enable the usage of the vast majority of masonry materials that are currently offered in Europe. Producing European specifications for popular masonry unit types was how the process was really developed:

clay
calcium silicate
aggregate concrete (sandcrete blocks)
autoclaved aerated concrete
manufactured stone and
natural stone.

Each of the product properties in these requirements is backed by an appropriate test method using a declaration system. The superior structural use masonry units are put in Group 1, and the remaining masonry units are divided into Groups 2, 3, and 4, according to the product characteristics. Group 1 units typically contain no more than 25% void space. Between 25 and 55% of Group 2 clay and calcium silicate units are voids, while 25 to 60% of Group 2 aggregate concrete units are voided.

Normalised Compressive Strength of Masonry

The compressive strength of masonry units had to be normalised because to variations in unit sizes that are available throughout Europe and in testing methods. The compressive strength is converted to the air-dried compressive strength of an identical 100 mm wide 100 mm high unit of the same material to provide the normalised compressive strength, or f_b.

The masonry’s normalised compressive strengths, f_b, are given by;

f_b = conditioning factor × shape factor × declared mean compressive strength

Compressive Strength of Mortar

Table 1 below lists the masonry mortar mixtures that the UK recommends using to reach the required strength specified in EC 6. Since the selection and designation of masonry mortars in EC 6 and BS 5628 are identical, similar mortars may be specified for a certain application or exposure.

Masonry mortars may be described by their compressive strength, which is indicated as the letter M followed by the compressive strength in N/mm², for example, M4, or by their mix proportion, for example, 1:1:6, which denotes the cement-lime-sand proportions by volume. The latter, however, has the advantage of producing mortars that are known to be durable and should generally be employed in practise.

Compressive Strength	Cement-lime-sand with or without air-entrainment	Cement-sand with or without air-entrainment	Masonry cement-sand	Masonry cement-sand	Mortar designation
M12	1:0 to 1/4:3	–	–		(i)
M6	1:1/2:4 to 4.5	1:3 to 4	1:2.5 to 3.5	1:3	(ii)
M4	1:1:5 to 6	1:5 to 6	1:4 to 5	1:3.5 to 4	(iii)
M2	1:2:8 to 9	1:7 to 8	1:5.5 to 6.5	1:4.5	(iv)

Table 1: Types of mortars (Table 2 of National Annex to EC 6)

Unit Manufacturing Control

According to the manufacturing control, units produced in compliance with European criteria can also be divided into Category I or Category II. Category I units are those when the manufacturer employs a quality-control programme and there is a less than 5% chance that the units will not achieve the specified compressive strength. Masonry units rated as Category II are not meant to meet the Category I degree of trust. The category of the masonry unit delivered must be declared by the manufacturer.

Class of Execution Control

EC 6 permits up to five classes of execution control, but, like BS 5628, the UK National Annex only uses two classes, namely 1 and 2. The requirements for workmanship in EN 1996: Part 2 (EC 6-2), including adequate supervision and inspection, shall be followed whenever the work is performed, in accordance with Table 1 of the National Annex, and in addition:

(a) The specification, supervision, and control ensure that the construction is compatible with the use of the appropriate safety factors specified in EC 6;
(b) the mortar complies with BS EN 998-2, if it is manufactured in a factory, or if it is site-mixed, preliminary compressive strength tests conducted on the mortar to be used, in accordance with BS EN 1015-2 and BS EN 1015-11, indicate conformity to the strength requirements specified in EC 6.

Every time the work is completed in accordance with the craftsmanship guidelines in EC 6-2, including the necessary supervision, class 2 execution control should be adopted. According to BS 5628, Class 1 execution control is equivalent to the “special” category of construction control, whereas Class 2 is equivalent to the “regular” category.

Characteristic Compressive Strength of Masonry

The characteristic compressive strength of unreinforced masonry, f_k, built with general-purpose mortar can be determined using the following expression

f_k = Kf_b^0.7f_m^0.3

where;
f_m is the compressive strength of general-purpose mortar but not exceeding 20 N/mm² or 2f_b, whichever is the smaller
f_b is the normalised compressive strength of the masonry units
K is a constant obtained from the Table below (Table 4 of the UK Annex to EC6).

Masonry Unit	Group	Values of K for general-purpose mortar
Clay	Group 1 Group 2	0.50 0.40
Calcium Silicate	Group 1 Group 2	0.50 0.40
Aggregate Concrete	Group 1 Group 1 (units laid flat) Group 2	0.55 0.50 0.52

Table 2: Values of K (based on Table 4 of the National Annex to EC 6)

Partial Factor for Materials (γ_M)

The values of the material property partial factors for the ultimate limit state listed in Table 1 of the National Annex to EC 6 are shown in Table 3. As is evident, they essentially depend on the type of unit, the type of execution control, and the level of stress.

	Class of execution control 1	Class of execution control 2
When in a state of direct or flexural compression Unreinforced masonry made with: units of category I units of category II	2.3 2.6	2.7 3.0
When in a state of flexural tension units of category I and II	2.3	2.7

Table 3: Values of γ_M for ultimate limit state (based on Table 2.3 of EC 6 and Table 1 of the National Annex to EC 6)

Design of unreinforced masonry walls subjected to vertical loading

Having discussed the basics, the following outlines EC 6 rules for the design of vertically loaded walls as set out in section 6.1. The approach is very similar to that in BS 5628 and principally involves checking that the design value of the vertical load, N_Ed, is less than or equal to the design value of the vertical resistance of the wall, N_Rd, i.e.

N_Ed ≤ N_Rd

According to clause 6.1.2.1 of EC 6, the design vertical load resistance of a single leaf unreinforced masonry wall per unit length, N_Rd, is given by;

N_Rd = Φ_i,m tf_k/γ_M

where;
Φ_i,m is the capacity reduction factor, Φi or Φm, as appropriate
t is the thickness of the wall
f_k is the characteristic compressive strength of masonry
γ_M is the material factor of safety for masonry determined from Table 3.

Effective Height of Masonry Wall

The effective wall height is a function of the actual wall height, h, and end/edge restraints. It can be taken as;

h_ef= ρ_nh

ρ_n is a reduction factor where n = 2, 3 or 4 depending on the number of restrained and stiffened edges. Thus, n = 2 for walls restrained at the top and bottom only, n = 3 for walls restrained top and bottom and stiffened on one vertical edge with the other vertical edge free and n = 4 for walls restrained top.

Worked Example on Masonry wall panel design

In accordance with EN1996-1-1:2005 + A1:2012 incorporating Corrigenda February 2006 and July 2009 and the UK national annex

Masonry panel details
Unreinforced masonry wall without openings
Panel length; L = 3600 mm
Panel height; h = 2700 mm
Panel support conditions: All edges supported

Effective height of masonry walls
Reduction factor; ρ₂ = 1.000
ρ₄ = ρ₂ / (1 + [ρ₂ × h/L]²) = 0.640
Effective height of wall – eq 5.2; h_ef = ρ₄ × h = 1728 mm

Single-leaf wall construction details
Wall thickness; t = 150 mm
Effective thickness; t_ef = t = 150 mm

Masonry details
Masonry type; Aggregate concrete – Group 2
Compressive strength of masonry; f_c = 2.9 N/mm²
Height of unit; h_u = 225 mm
Width of unit; w_u = 150 mm
Conditioning factor; k = 1.0 – Conditioning to the air dry condition in accordance with cl.7.3.2
Shape factor – Table A.1; d_sf = 1.3

Norm. mean compressive strength of masonry; f_b = f_c × k × d_sf = 3.77 N/mm²

Density of masonry; γ = 18 kN/m³
Mortar type; M2 – General purpose mortar
Compressive strength of masonry mortar; f_m = 2 N/mm²
Compressive strength factor – Table NA.4; K = 0.70

Characteristic compressive strength of masonry – eq 3.1
f_k = K × f_b^0.7 × f_m^0.3 = 2.182 N/mm²

Characteristic flexural strength of masonry having a plane of failure parallel to the bed joints – Table NA.6
f_xk1 = 0.167 N/mm²

Characteristic flexural strength of masonry having a plane of failure perpendicular to the bed joints – Table NA.6
f_xk2 = 0.338 N/mm²

Lateral loading details
Characteristic wind load on panel; W_k = 0.700 kN/m²

Vertical loading details
Permanent load on top of wall; G_k = 21 kN/m;
Variable load on top of wall; Q_k = 7 kN/m;

Partial factors for material strength
Category of manufacturing control; Category II
Class of execution control; Class 2
Partial factor for masonry in compressive flexure; γ_Mc = 3.00
Partial factor for masonry in tensile flexure; γ_Mt = 2.70
Partial factor for masonry in shear; γ_Mv = 2.50

Slenderness ratio of masonry walls
Allowable slenderness ratio;SR_all = 27
Slenderness ratio; SR = h_ef / t_ef = 11.5

PASS – Slenderness ratio is less than maximum allowable

Partial safety factors for design loads
Partial safety factor for permanent load; γ_fG = 1.35
Partial safety factor for variable imposed load; γ_fQ = 1.5
Partial safety factor for variable wind load; γ_fW = 0.75

Reduction factor for slenderness and eccentricity – Section 6.1.2.2

Vertical load at top of wall; N_id = γ_fGG_k + γ_fQQ_k = 38.85 kN/m
Moment at top of wall due to vertical load; M_id = γ_fGG_ke_G + γ_fQQ_ke_Q = 0 kNm/m
Initial eccentricity – cl.5.5.1.1; e_init = h_ef / 450 = 3.8 mm
Moment at top of wall due to horizontal load; M_Eid = 0 kNm/m
Eccentricity at top of wall due to horizontal load; e_h = 0 mm

Eccentricity at top of wall – eq.6.5;
e_i = max(M_id / N_id + e_h + e_init, 0.05t) = 7.5 mm

Reduction factor at top of wall – eq.6.4;
Φ_i = max(1 – 2e_i/t, 0) = 0.9

Vertical load at middle of wall;
N_md = γ_fG(G_k + γth/2) + γ_fQQ_k = 43.771 kN/m

Moment at middle of wall due to vertical load;
M_md = γ_fGG_ke_G + γ_fQQ_ke_Q = 0 kNm/m

Moment at middle of wall due to horizontal load; M_Emd = 0.087 kNm/m
Eccentricity at middle of wall due to horizontal load; e_hm = M_Emd / N_md = 2 mm
Eccentricity at middle of wall due to loads – eq.6.7; e_m = M_md / N_md + e_hm + e_init = 5.8 mm
Eccentricity at middle of wall due to creep; e_k = 0 mm
Eccentricity at middle of wall – eq.6.6; e_mk = max(e_m + e_k, 0.05t) = 7.5 mm

From eq.G.2; A₁ = 1 – 2 × e_mk/t = 0.9

Short-term secant modulus of elasticity factor; K_E = 1000
Modulus of elasticity – cl.3.7.2; E = K_E × f_k = 2182 N/mm²
Slenderness – eq.G.4; λ = (h_ef / t_ef) × √(f_k / E) = 0.364
From eq.G.3; u = (λ – 0.063) / (0.73 – 1.17e_mk/t) = 0.449

Reduction factor at middle of wall – eq.G.1;
Φ_m = max(A₁ × e_e^{-(u × u)/2}, 0) = 0.814

Reduction factor for slenderness and eccentricity;
Φ = min(F_i, F_m) = 0.814

Verification of unreinforced masonry walls subjected to mainly vertical loading – Section 6.1.2

Design value of the vertical load;
N_Ed = max(N_id, N_md) = 43.771 kN/m

Design compressive strength of masonry;
f_d = f_k /γ_Mc = 0.727 N/mm²

N_Rd = Φ × t × f_d = 88.786 kN/m

PASS – Design vertical resistance exceeds applied design vertical load

Unreinforced masonry walls subjected to lateral loading

Partial safety factors for design loads

Partial safety factor for permanent load; γ_fG = 1
Partial safety factor for variable imposed load; γ_fQ = 0
Partial safety factor for variable wind load; γ_fW = 1.5

Limiting height and length to thickness ratios for walls under the serviceability limit state – Annex F

Length to thickness ratio; L/t = 24
Limiting height to thickness ratio – Figure F.1; 80
Height to thickness ratio; h/t = 18

PASS – Limiting height to thickness ratio is not exceeded

Design moments of resistance in panels

Self-weight at middle of wall;
S_wt = 0.5 × h × t × γ = 3.645 kN/m

Design compressive strength of masonry;
f_d = f_k/γ_Mc = 0.727 N/mm²

Design vertical compressive stress;
s_d = min(γ_fG × (G_k + S_wt) / t, 0.15Ff_d) = 0.089 N/mm²

Design flexural strength of masonry parallel to bed joints
f_xd1 = f_xk1/γ_Mt = 0.062 N/mm²

Apparent design flexural strength of masonry parallel to bed joints
f_xd1,app = f_xd1 + σ_d = 0.151 N/mm²

Design flexural strength of masonry perpendicular to bed joints
f_xd2 = f_xk2 /γ_Mt = 0.125 N/mm²

Elastic section modulus of wall;
Z = t²/6 = 3750000 mm³/m

Moment of resistance parallel to bed joints – eq.6.15
M_Rd1 = f_xd1,appZ = 0.564 kNm/m

Moment of resistance perpendicular to bed joints – eq.6.15
M_Rd2 = f_xd2Z = 0.469 kNm/m

Design moment in panels
Orthogonal strength ratio; m = f_xd1,app / f_xd2 = 1.20

Using yield line analysis to calculate bending moment coefficient
Bending moment coefficient; α = 0.027
Design moment in wall; M_Ed = γ_fW × α × W_k × L² = 0.367 kNm/m

PASS – Resistance moment exceeds design moment

Summary

	Allowable	Actual	Utilisation
Slenderness ratio;	27	11.5	0.427	PASS
Vertical loading on wall;	88.786 kN/m	43.771 kN/m	0.493	PASS
Height to thickness ratio;	80.000	18.000	0.225	PASS
Design moment to wall;	0.469 kNm/m	0.367 kNm/m	0.782	PASS

Digital Twin in the AEC Industry

By

Anuoluwapo Kolade

-

October 29, 2022

Table of Contents

Is there a better way to show and view a project than by digital twin? For example, suppose you plan to stay true to the design of a project during construction. Is there a better way to achieve this than by digital twin? Perhaps what comes to your mind when you hear “digital twin” is BIM. But, then, you may be surprised to know that digital twin goes beyond BIM.

For instance, BIM focuses on the system and sub-systems that ensure a building is constructed correctly. However, digital twin can be a combination of a building and a computational model to monitor, manage, and optimize the building. Thus, digital twin concentrates on the operation, performance, and functionality of the building, including use patterns, occupant behaviours, and space utilization. Nevertheless, both tools are vital in this age of intelligent building design, operation, maintenance, and management.

This article covers everything about digital twin in the architecture, engineering, and construction (AEC) industry, including the purpose, types, examples, and benefits of digital twin in construction. By reading this article, you will discover why digital twin is an essential tool in the construction industry and built environment and why you should now adopt and invest in digital twin.

Purpose of Digital Twin

At present, digital twin is growing in importance in the AEC industry because it is used to integrate real-time data from an existing asset with its digital representation to provide insights across the asset’s lifecycle. Digital twin is a proven game-changer in the AEC industry because it solves significant asset design, construction, and operation challenges. Adopting digital twins is also the best way to improve asset performance.

When you think of digital twin, virtual creations or exact digital replicas of an asset should come to your mind. For example, digital twin is like the 3D interior renderings of a building, and it is a tool that helps to utilize, manage, and optimize the spaces in the building. Thus, with the digital twin, you can easily make simulations and predictions based on real-world data and conditions to make informed decisions and develop failure-proof working methods.

You may ask where BIM comes in or how digital twin uses BIM. The fact is that BIM remains the most efficient path to creating an accurate, high-value digital twin. Usually, BIM uses the data created during the planning and design of a project. Likewise, digital twin uses this same data and the data created during the construction and operation of the completed project. Thus, this data can also be used to make calculated and informed decisions for future projects.

Types of Digital Twin

Depending on the degree of maturity and digital transformation you desire, the types of digital twin are briefly discussed below.

Descriptive twin

3D architectural models and renderings and BIM comprise descriptive twin. Descriptive twin provides data and editable visual and photorealistic representations of a project for scheduling construction processes and actual construction.

Informative twin

When you need to make informed decisions about facility management operation and estimation, you need information twin. Information twin involves using BIM with sensors, simulations, and operations data to draw insights at any given time within an asset’s lifecycle.

Predictive twin

Predictive twin involves using BIM with IoT technologies to track, capture, visualize, and analyze real-time data to identify potential issues during an asset’s lifecycle.

Comprehensive twin

Comprehensive twin involves leveraging BIM and AI for data-based advanced modelling and simulation to carry out prescriptive analysis and make recommendations to aid decision-making in future situations.

Autonomous twin

An autonomous twin is the ideal digital twin. Unlike comprehensive twin, where professionals still make final decisions, autonomous twin goes further to make decisions through AI-driven advanced simulation and 3D visualization algorithms.

Examples of digital twin

A digital twin is several things to the AEC industry, mainly because it is linked to assets, the spaces in assets, and the people who use these assets and spaces. For instance, digital twin is a tool for safety planning, asset management, space optimization, and improving occupant or employee experience. Examples of digital twin in the AEC industry are highlighted below.

• Developing new building construction models

There are several designs that architects conceive, but that never come to reality. This happens because architects sometimes need clarification on the practicality of such designs and if these designs meet all stability and safety requirements. To make matters worse, unfortunately, it is almost impossible to test such designs in the real world to see if they will work or not because of the capital, time, and other resources involved. However, a digital twin can solve this challenge.

With a digital twin, architects and other professionals in the AEC industry can test a design idea in reality-based simulations that would use real-world data and involve all the necessary real-world factors. Similarly, they can test such design’s stability, safety, sustainability, and practicality and get accurate feedback as if the design was executed in the real world.

• Reduction of construction and operating costs during building renovations

One significant advantage of digital twin is that it makes coordinating teams and construction processes easier. In addition, one can run several sequencing and logistics scenarios in advance. Thus, giving you an idea of what you are to do and how you are to do them during the construction of a project. Also, it is possible to feed back real-world data to the digital twin in cases where situations and events change.

• “What if” analysis of safety systems and emergency action plans

With “what if” analysis, you can simulate a scenario to know how a particular asset is performing and will perform under specific conditions. For example, a digital twin can be used to understand how a rail line was designed and constructed, how it functions, and its limitations. Similarly, a digital twin can provide workers with real-time tracking and alerts of work areas. Thus, reducing risks to workers, making construction attractive to people, and increasing overall efficiency and work quality on construction sites.

• Productivity and collaboration improvements through space delegation

Important information about an existing asset can be stored, analyzed, and updated throughout the asset’s lifetime. This information becomes helpful in making informed and calculated decisions when embarking on future projects. This information can also assist with de-risking the execution of the project.

• Asset performance and sustainability optimization through use monitoring

Professionals can improve the performance and sustainability of an asset from the insights obtained from a digital twin. For example, a digital twin can be used to predict maintenance events that may arise and plan the required maintenance activities. Consequently, unexpected circumstances and maintenance costs will be reduced throughout the asset’s lifetime. Furthermore, a digital twin can detect a fault or abnormality in a building and automatically dispatch a worker to rectify such issues without delay.

Benefits of digital twin

BIM is known to be very analytical. However, digital twin improves the digital capabilities of BIM and makes it easier for asset owners, operators and managers to visualize the real-time status, working conditions, and position of physical assets. Some benefits of digital twin for firms, owners and managers in the AEC industry are discussed below.

• Competitive edge
If you own a firm, digital twin can differentiate you from your competitors and put you ahead of them. In addition, with digital twin, firms can prove to their clients that they can provide them value and a broader range of services by leveraging technology to harness and utilize real-time data.

• Savings on money
Digital twin allows firms, asset owners, operators and managers to make informed and calculated decisions, thus helping to save cost and extending the value and lifecycle of assets.

• Savings on time
Usually, there are conversations between firms and stakeholders before a project kicks off. These conversations and collaborations help to prioritize the use and performance requirements of such projects or assets. Therefore, it becomes easier to plan to achieve the project goals and eliminate time wastage. Similarly, asset owners and operators can seamlessly proceed with operating the asset without having to bother about misplaced operation or maintenance documents.

• Streamlined facility management
Digital twin usually offloads most of the burdens of asset management. For instance, a digital twin of an asset can be used to spot the problem areas and provide potential solutions to the problems that technicians can work with.

Conclusion

A digital twin is simply an exact, digital, and dynamic replica of an asset after its design and construction, for example, a building. Thus, the digital twin of the building can change with its use, just like the physical building, because it is responsive to data input. Its use will also change as more data is supplied from sensors, IoT, and AI.

Just like BIM technology has become indispensable to the AEC industry, so is digital twin technology, especially for facility management. However, models and data from BIM are of little or no use to asset owners, operators, and managers once a project is completed. Therefore, a digital twin is vital for asset owners, operators and managers, and professionals in the AEC industry because it provides insights to optimize assets, systems, and workplaces.

References

[1] Deng, M., Menassa, C. C., and Kamat, V. R. (2021). “From BIM to Digital Twins: a Systematic Review of the Evolution of Intelligent Building Representations in the AEC-FM Industry”. ITcon 26, 58–83. Doi:10.36680/j.itcon.2021.005

[2] Autodesk, Inc. (n.d.). Retrieved from https://www.autodesk.com/solutions/digital-twin/architecture-engineering-construction

Design of Biaxial Eccentrically Loaded Pad Footing

By

Ubani Obinna

-

October 27, 2022

A biaxial eccentrically loaded pad footing occurs when the column transmitting load to the foundation is subjected to compressive axial force and bending moment in the two principal axes. As a result of the biaxial bending, two eccentricities e_x and e_y of the axial load occur on the pad footing, thereby leading to non-uniform pressure distribution on the foundation. The effects of such non-uniform pressure distribution must be accounted for during the geotechnical and structural design of the pad foundation.

This article will consider the geotechnical and structural design of a biaxial eccentrically loaded pad footing. The footing for a single column may be made square in plan, but where there is a large moment acting about only one axis it may be more economical to have a rectangular base.

pad footing pressure distributions — Different pressure distribution of pad footings

When a bending moment M and axial force N are acting on a pad foundation, the pressures are given by the equation for axial load plus bending. This condition is valid provided there is positive contact between the pad base and the ground along the complete length D of the footing so that;

p = N/BD ± My/I

where I is the second-moment area of the base about the axis of bending and y is the distance from the axis to where the pressure is being calculated.

Substituting for I = BD³/12 and y = D/2, the maximum pressure is;

p_max = N/BD + 6M/BD²

and the minimum pressure is;

p_min = N/BD – 6M/BD²

For biaxially loaded footings, this pressure must be verified in both directions, and the maximum pressure should not exceed the allowable bearing capacity of the soil. Furthermore, the reinforcement design must also be carried out in both directions.

Design Example of Biaxial Eccentrically Loaded Pad Footing

A 1500 x 1500mm pad foundation is subjected to the following loads from a 250 mm x 250 mm square column;

Permanent load axial load; F_Gz1 = 650.0 kN
Variable laxial load; F_Qz1 = 135.0 kN
Permanent moment in x; M_Gx1 = 25.0 kNm
Permanent moment in y; M_Gy1 = 21.0 kNm
Variable moment in x; M_Qx1 = 13.0 kNm
Variable moment in y; M_Qy1 = 11.0 kNm

The design is to be done in accordance with EN1997-1:2004 incorporating Corrigendum dated February 2009 and the UK National Annex incorporating Corrigendum No.1

Pad foundation details

Length of foundation; L_x = 1500 mm
Width of foundation; L_y = 1500 mm
Foundation area; A = L_x × L_y = 2.250 m²
Depth of foundation (thickness of footing); h = 500 mm
Depth of soil over foundation; h_soil = 600 mm
Level of water; h_water = 0 mm
Density of water; γ_water = 9.8 kN/m³
Density of concrete; γ_conc = 25.0 kN/m³

Column details
Length of column; l_x1 = 250 mm
Width of column; l_y1 = 250 mm
position in x-axis; x₁ = 750 mm
position in y-axis; y₁ = 750 mm

Library item: Column details output

Soil properties
Density of soil; γ_soil = 18.0 kN/m³
Characteristic cohesion; c’_k = 15 kN/m²
Characteristic effective shear resistance angle; φ’_k = 25 deg
Characteristic friction angle; δ_k = 20 deg

Foundation loads
Permanent surcharge load; F_Gsur = 5.0 kN/m²
Self weight; F_swt = h × γ_conc = 12.5 kN/m²
Soil weight; F_soil = h_soil × γ_soil = 10.8 kN/m²

Column loads
Permanent load in z; F_Gz1 = 650.0 kN
Variable load in z; F_Qz1 = 135.0 kN
Permanent moment in x; M_Gx1 = 25.0 kNm
Permanent moment in y; M_Gy1 = 21.0 kNm
Variable moment in x; M_Qx1 = 13.0 kNm
Variable moment in y; M_Qy1 = 11.0 kNm

Design Approach 1 (DA 1) – Combination 1

Partial factors on actions – Combination1
Partial factor set; A1
Permanent unfavourable action – Table A.3; γ_G = 1.35
Permanent favourable action – Table A.3; γ_Gf = 1.00
Variable unfavourable action – Table A.3; γ_Q = 1.50
Variable favourable action – Table A.3; γ_Qf = 0.00

Partial factors for soil parameters – Combination1
Soil factor set; M1
Angle of shearing resistance – Table A.4; γ_φ’ = 1.00
Effective cohesion – Table A.4; γ_c’ = 1.00
Weight density – Table A.4; γ_g = 1.00

Partial factors for spread foundations – Combination1
Resistance factor set; R1
Bearing – Table A.5; γ_R.v = 1.00
Sliding – Table A.5; γ_R.h = 1.00

Bearing Resistance

Forces on foundation
Force in z-axis; F_dz = γ_G × [A × (F_swt + F_soil + F_Gsur) + F_Gz1] + γ_QF_Qz1 = 1166.0 kN

Moments on foundation
Moment in x-axis;
M_dx = γ_G × (A × (F_swt + F_soil + F_Gsur) × L_x/2 + F_Gz1x₁) + γ_GM_Gx1 + γ_QF_Qz1x₁ + γ_QM_Qx1 = 927.7 kNm

Moment in y-axis;
M_dy = γ_G × (A × (F_swt + F_soil + F_Gsur) × L_y/2 + F_Gz1y₁) + γ_GM_Gy1 + γ_QF_Qz1y₁ + γ_QM_Qy1 = 919.3 kNm

Eccentricity of base reaction

Eccentricity of base reaction in x-axis;
e_x = M_dx / F_dz – L_x / 2 = 46 mm

Eccentricity of base reaction in y-axis;
e_y = M_dy / F_dz – L_y / 2 = 38 mm

Effective area of base

Effective length;
L’_x = L_x – 2 × e_x = 1409 mm

Effective width;
L’_y = L_y – 2 × e_y = 1423 mm

Effective area;
A’ = L’_x × L’_y = 2.005 m²

Pad base pressure

Design base pressure; f_dz = F_dz / A’ = 581.6 kN/m²
Design angle of shearing resistance; φ’_d = tan^-1(tan(φ’_k) / γ_φ’) = 25.000 deg
Design effective cohesion; c’_d = c’_k / γ_c’ = 15.000 kN/m²

Effective overburden pressure;
q = (h + h_soil) × γ_soil – h_water × γ_water = 19.800 kN/m²

Design effective overburden pressure;
q’ = q / γ_g = 19.800 kN/m²

Bearing resistance factors;
N_q = Exp(π × tan(φ’_d)) × [tan(45 deg + φ’_d / 2)]² = 10.662
N_c = (N_q – 1) × cot(φ’_d) = 20.721
N_γ = 2 × (N_q – 1) × tan(φ’_d) = 9.011

Foundation shape factors;
s_q = 1 + (L’_x / L’_y) × sin(φ’_d) = 1.418
s_{_γ} = 1 – 0.3 × (L’_x / L’_y) = 0.703
s_c = (s_q × N_q – 1) / (N_q – 1) = 1.462

Load inclination factors;
H = 0.0 kN
m_y = [2 + (L’_y / L’_x)] / [1 + (L’_y / L’_x)] = 1.497
m_x = [2 + (L’_x / L’_y)] / [1 + (L’_x / L’_y)] = 1.503
m = m_x = 1.503
i_q = [1 – H / (F_dz + A’ × c’_d × cot(φ’_d))]^m = 1.000
i_γ = [1 – H / (F_dz + A’ × c’_d × cot(φ’_d))]^{m + 1} = 1.000
i_c = i_q – (1 – i_q) / (N_c × tan(φ’_d)) = 1.000

Ultimate bearing capacity;
n_f = c’_dN_cs_ci_c + q’N_qs_qi_q + 0.5γ_soilL’_xN_{_γ}s_γi_{_γ} = 834.0 kN/m²

PASS – Ultimate bearing capacity exceeds design base pressure

Design Approach 1 (DA 1) – Combination 2

Partial factors on actions – Combination2
Partial factor set; A2
Permanent unfavourable action – Table A.3; γ_G = 1.00
Permanent favourable action – Table A.3; γ_Gf = 1.00
Variable unfavourable action – Table A.3; γ_Q = 1.30
Variable favourable action – Table A.3; γ_Qf = 0.00

Partial factors for soil parameters – Combination2
Soil factor set; M2
Angle of shearing resistance – Table A.4; γ_φ’ = 1.25
Effective cohesion – Table A.4; γ_c’ = 1.25
Weight density – Table A.4; γ_g = 1.00

Partial factors for spread foundations – Combination2
Resistance factor set; R1
Bearing – Table A.5; γ_R.v = 1.00
Sliding – Table A.5; γ_R.h = 1.00

Bearing resistance (Section 6.5.2)

Forces on foundation
Force in z-axis;
F_dz = γ_G × (A × (F_swt + F_soil + F_Gsur) + F_Gz1) + γ_QF_Qz1 = 889.2 kN

Moments on foundation

Moment in x-axis;
M_dx = γ_G × (A × (F_swt + F_soil + F_Gsur) × L_x/2 + F_Gz1 × x₁) + γ_GM_Gx1 + γ_QF_Qz1x₁ + γ_QM_Qx1 = 708.8 kNm

Moment in y-axis;
M_dy = γ_G × (A × (F_swt + F_soil + F_Gsur) × L_y/2 + F_Gz1 × y₁) + γ_GM_Gy1 + γ_QF_Qz1y₁ + γ_QM_Qy1 = 702.2 kNm

Eccentricity of base reaction

Eccentricity of base reaction in x-axis;
e_x = M_dx / F_dz – (L_x /2) = 47 mm

Eccentricity of base reaction in y-axis;
e_y = M_dy / F_dz – (L_y/2) = 40 mm

Effective area of base

Effective length;
L’_x = L_x – 2e_x = 1406 mm

Effective area;
A’ = L’_x × L’_y = 1.997 m²

Effective width;
L’_y = L_y – 2e_y = 1421 mm

Pad base pressure

Design base pressure; f_dz = F_dz / A’ = 445.3 kN/m²
Ultimate bearing capacity under drained conditions (Annex D.4)

Design angle of shearing resistance;
φ’_d = tan^-1(tan(φ’_k) / γ_f’) = 20.458 deg

Design effective cohesion;
c’_d = c’_k / γ_c’ = 12.000 kN/m²

Effective overburden pressure;
q = (h + h_soil) × γ_soil – h_water × γ_water = 19.800 kN/m²

Design effective overburden pressure;
q’ = q/γ_g = 19.800 kN/m²

Bearing resistance factors;
N_q = Exp(π × tan(φ’_d)) × (tan(45 deg + φ’_d / 2))² = 6.698
N_c = (N_q – 1) × cot(φ’_d) = 15.273
N_γ = 2 × (N_q – 1) × tan(φ’_d) = 4.251

Foundation shape factors;
s_q = 1 + (L’_x / L’_y) × sin(φ’_d) = 1.346
s_γ = 1 – 0.3 × (L’_x / L’_y) = 0.703
s_c = (s_q × N_q – 1) / (N_q – 1) = 1.407

Load inclination factors;
H = 0.0 kN
m_y = [2 + (L’_y / L’_x)] / [1 + (L’_y / L’_x)] = 1.497
m_x = [2 + (L’_x / L’_y)] / [1 + (L’_x / L’_y)] = 1.503
m = m_x = 1.503
i_q = [1 – H / (F_dz + A’ × c’_d × cot(φ’_d))]^m = 1.000
i_{_γ} = [1 – H / (F_dz + A’ × c’_d × cot(φ’_d))]^{m + 1} = 1.000
i_c = i_q – (1 – i_q) / (N_c × tan(φ’_d)) = 1.000

Ultimate bearing capacity;
n_f = c’_dN_cs_ci_c + q’N_qs_qi_q + 0.5 γ_soilL’_xN_γ s_g i_γ = 474.1 kN/m²

PASS – Ultimate bearing capacity exceeds design base pressure

Foundation design (EN1992-1-1:2004)

In accordance with EN1992-1-1:2004 incorporating Corrigendum dated January 2008 and the UK National Annex incorporating National Amendment No.1

Concrete details

Concrete strength class; C25/30
Characteristic compressive cylinder strength; f_ck = 25 N/mm²
Characteristic compressive cube strength; f_ck,cube = 30 N/mm²
Mean value of compressive cylinder strength;f_cm = f_ck + 8 N/mm² = 33 N/mm²
Mean value of axial tensile strength; f_ctm = 0.3 N/mm² × (f_ck)^2/3 = 2.6 N/mm²
5% fractile of axial tensile strength;f_ctk,0.05 = 0.7 × f_ctm = 1.8 N/mm²
Secant modulus of elasticity of concrete; E_cm = 22 kN/mm² × [f_cm/10]^0.3 = 31476 N/mm²

Partial factor for concrete (Table 2.1N); γ_C = 1.50
Compressive strength coefficient (cl.3.1.6(1)); a_cc = 0.85
Design compressive concrete strength (exp.3.15); f_cd = a_cc × (f_ck / γ_C) = 14.2 N/mm²
Tens.strength coeff.for plain concrete (cl.12.3.1(1)); a_ct,pl = 0.80
Des.tens.strength for plain concrete (exp.12.1); f_ctd,pl = a_ct,pl × (f_ctk,0.05 / γ_C) = 1.0 N/mm²

Maximum aggregate size; h_agg = 20 mm
Ultimate strain – Table 3.1; ε_cu2 = 0.0035
Shortening strain – Table 3.1;ε_cu3 = 0.0035
Effective compression zone height factor; λ = 0.80
Effective strength factor; h = 1.00
Bending coefficient k₁; K₁ = 0.40
Bending coefficient k₂; K₂ = 1.00 × (0.6 + 0.0014/ε_cu2) =1.00
Bending coefficient k₃; K₃ =0.40
Bending coefficient k₄; K₄ =1.00 × (0.6 + 0.0014/ε_cu2) = 1.00

Reinforcement details

Characteristic yield strength of reinforcement; f_yk = 500 N/mm²
Modulus of elasticity of reinforcement; E_s = 210000 N/mm²
Partial factor for reinforcing steel (Table 2.1N); γ_S = 1.15
Design yield strength of reinforcement; f_yd = f_yk / γ_S = 435 N/mm²
Nominal cover to reinforcement; c_nom = 50 mm

Rectangular section in flexure (x-axis)

Design bending moment; M_Ed.x.max = 160.7 kNm
Depth to tension reinforcement; d = h – c_nom – φ_x.bot / 2 = 444 mm
K = M_Ed.x.max / (L_y × d² × f_ck) = 0.022
K’ = (2 × h × a_cc/γ_C) × (1 – λ(d – K₁)/(2K₂)) × (λ(d – K₁)/(2K₂))
K’ = 0.207

K’ > K – No compression reinforcement is required

Lever arm; z = min(0.5 + 0.5 × (1 – 2K / (h × a_cc /γ_c))^0.5, 0.95) × d = 422 mm
Depth of neutral axis; x = 2.5(d – z) = 55 mm

Area of tension reinforcement required;
A_sx.bot.req = M_Ed.x.max / (f_ydz) = 876 mm²

Tension reinforcement provided;
10Y12@155 c/c bottom (A_sx.bot.prov = 1131 mm²)

Minimum area of reinforcement (exp.9.1N);
A_s.min = max(0.26 × f_ctm / f_yk, 0.0013) × L_y × d = 888 mm²

Maximum area of reinforcement (cl.9.2.1.1(3));
A_s.max = 0.04 × L_y × d = 26640 mm²

PASS – Area of reinforcement provided is greater than area of reinforcement required

Rectangular section in shear (x-axis)

Design shear force;
abs(V_Ed.x.min) = 162.8 kN
C_Rd,c = 0.18 /γ_C = 0.120
k = min(1 + √(200 mm / d), 2) = 1.680

Longitudinal reinforcement ratio;
ρ_l = min(A_sx.bot.prov / (L_y × d), 0.02) = 0.002
v_min = 0.035k^3/2 × f_ck^0.5 = 0.381 N/mm²

Design shear resistance (exp.6.2a & 6.2b);
V_Rd.c = max(C_Rd.c × k × (100 N²/mm⁴ × ρ_l × f_ck)^1/3, v_min) × L_y × d
V_Rd.c = 247 kN

PASS – Design shear resistance exceeds design shear force

Rectangular section in flexure (y-axis)

Design bending moment;
M_Ed.y.max = 157.5 kNm
Depth to tension reinforcement;
d = h – c_nom – f_x.bot – φ_y.bot / 2 = 432 mm
K = M_Ed.y.max / (L_x × d² × f_ck) = 0.02
K’ = (2h × a_cc/γ_C) × (1 – λ × (d – K₁)/(2K₂)) × (λ × (d – K₁)/(2K₂))
K’ = 0.207

K’ > K – No compression reinforcement is required

Lever arm;
z = min(0.5 + 0.5 × (1 – 2K / (h × a_cc/γ_C))^0.5, 0.95) × d = 410 mm

Depth of neutral axis;
x = 2.5(d – z) = 54 mm

Area of tension reinforcement required;
A_sy.bot.req = M_Ed.y.max / (f_ydz) = 883 mm²

Tension reinforcement provided;
12Y12@125 c/c A_sy.bot.prov = 1357 mm²

Minimum area of reinforcement (exp.9.1N);
A_s.min = max(0.26f_ctm / f_yk, 0.0013) × L_x × d = 864 mm²

Maximum area of reinforcement (cl.9.2.1.1(3));
A_s.max = 0.04 × L_x × d = 25920 mm²

PASS – Area of reinforcement provided is greater than the area of reinforcement required

Crack control

Limiting crack width; w_max = 0.3 mm
Variable load factor (EN1990 – Table A1.1); y₂ = 0.3
Serviceability bending moment; M_sls.y.max = 99 kNm
Tensile stress in reinforcement; s_s = M_sls.y.max / (A_sy.bot.prov × z) = 177.8 N/mm²
Load duration factor; k_t = 0.4

Effective depth of concrete in tension;
h_c.ef = min(2.5 × (h – d), (h – x) / 3, h/2) = 149 mm

Effective area of concrete in tension;
A_c.eff = h_c.ef × L_x = 223000 mm²

Mean value of concrete tensile strength;
f_ct.eff = f_ctm = 2.6 N/mm²

Reinforcement ratio;
ρ_p.eff = A_sy.bot.prov / A_c.eff = 0.006

Modular ratio; a_e = E_s / E_cm = 6.672

Bond property coefficient; k₁ = 0.8
Strain distribution coefficient; k₂ = 0.5
k₃ = 3.4
k₄ = 0.425

Maximum crack spacing (exp.7.11);
s_r.max = k₃ × (c_nom + f_x.bot) + k₁k₂k₄ × φ_y.bot / ρ_p.eff = 546 mm

Maximum crack width (exp.7.8);
w_k = s_r.max × max([s_s – k_t × (f_ct.eff / ρ_p.eff) × (1 + a_e × ρ_p.eff)] / E_s, 0.6 × s_s / E_s) = 0.277 mm

PASS – Maximum crack width is less than limiting crack width

Rectangular section in shear (y-axis)

Design shear force; abs(V_Ed.y.min) = 159.1 kN
C_Rd,c = 0.18/γ_C = 0.120
k = min(1 + √(200 mm / d), 2) = 1.680

Longitudinal reinforcement ratio;
r_l = min(A_sy.bot.prov / (L_x × d), 0.02) = 0.002
v_min = 0.035k^3/2 × f_ck^0.5 = 0.381 N/mm²

Design shear resistance (exp.6.2a & 6.2b);
V_Rd.c = max(C_Rd.c × k × (100 × r_l × f_ck)^1/3, v_min) × L_x × d
V_Rd.c = 247 kN

PASS – Design shear resistance exceeds design shear force

Punching shear

Strength reduction factor (exp 6.6N); v = 0.6[1 – f_ck / 250] = 0.540
Average depth to reinforcement; d = 438 mm
Maximum punching shear resistance (cl.6.4.5(3)); v_Rd.max = 0.5vf_cd = 3.825 N/mm²

k = min(1 + √(200 mm / d), 2) = 1.676

Longitudinal reinforcement ratio (cl.6.4.4(1));
r_lx = A_sx.bot.prov / (L_y × d) = 0.002
r_ly = A_sy.bot.prov / (L_x × d) = 0.002
r_l = min(√(r_lx × r_ly), 0.02) = 0.002
C_Rd,c = 0.18 / g_C =0.120

v_min = 0.035 k^3/2 × f_ck^0.5 = 0.380 N/mm²

Design punching shear resistance (exp.6.47);
v_Rd.c = max(C_Rd.c k (100r_lf_ck)^1/3, v_min) = 0.380 N/mm²

Design punching shear resistance at 1d (exp. 6.50);
v_Rd.c1 = (2d/d)v_Rd.c = 0.759 N/mm²

Punching shear perimeter at column face

Punching shear perimeter; u₀ = 1000 mm
Area within punching shear perimeter; A₀ = 0.063 m²
Maximum punching shear force; V_Ed.max = 1046 kN
Punching shear stress factor (fig 6.21N); β = 1.500

Maximum punching shear stress (exp 6.38);
v_Ed.max = β V_Ed.max / (u₀ × d) = 3.582 N/mm²

PASS – Maximum punching shear resistance exceeds maximum punching shear stress

Punching shear perimeter at 1d from column face

Punching shear perimeter; u₁ = 3752 mm
Area within punching shear perimeter; A₁ = 1.103 m²
Design punching shear force; V_Ed.1 = 480.5 kN
Punching shear stress factor (fig 6.21N); β = 1.500
Design punching shear stress (exp 6.38); v_Ed.1 = βV_Ed.1 / (u₁d) = 0.439 N/mm²

PASS – Design punching shear resistance exceeds increased design punching shear stress

Punching shear perimeter at 2d from column face

Punching shear perimeter; u₂ = 63 mm
Area within punching shear perimeter; A₂ = 2.250 m²
Design punching shear force; V_Ed.2 = 0 kN
Punching shear stress factor (fig 6.21N); β = 1.500
Design punching shear stress (exp 6.38); v_Ed.2 = βV_Ed.2 / (u₂ × d) = 0.001 N/mm²

PASS – Design punching shear resistance exceeds design punching shear stress

Piping Methods in Residential Buildings

By

Anuoluwapo Kolade

-

October 25, 2022

Table of Contents

Recently, there have been discussions on the choice of piping methods for plumbing systems in building structures. These discussions have arisen due to the implications of some piping methods on the structural system, durability, performance, and maintenance of building structures.

To worsen the case, the use of substandard construction materials such as low-quality pipes which deteriorates quickly has not helped. As a result, a lot of leakages occur in the plumbing system of such defective houses, soaking the walls, and making the building uninhabitable. Thus, this article discusses piping methods and recommends the best approach to help reduce building maintenance costs and prevent deterioration and eventual building collapse.

59B032E9 2842 4C74 B84B 0BFB9FA17F11 — Piping system deterioration in a building

Piping Methods

Piping is primarily used in building structures for plumbing purposes. Building plumbing systems consist of piping networks that distribute drinking water and safely dispose of sewage into sewerage systems. Three piping methods are commonly used in building structures. They are;

surface piping
conduit piping, and
duct piping.

These piping methods are discussed in detail below.

Surface Piping

When you see a building with PVC pipes running on the exterior of a building, then know that surface piping is employed for plumbing. This type of piping is the foremost piping method for plumbing works, and the major advantages of surface piping to other piping systems are ease of installation and maintenance.

However, surface piping usually needs to be more aesthetically pleasing and appealing. Thus, running pipes on the surface of buildings are often undesirable because they significantly affect building appearance. However, in instances where pipes or joints leak, they are so easy to detect, repair, or replace without affecting the building or breaking the walls.

Surface pipes can also run through openings created in walls. These openings are of small sizes, generally ranging between ½ to 6 inches. Thus, surface piping involves minor demolition and has negligible impact on the structural system, structural integrity, and performance of a building, especially when it is a frame structure. However, surface pipes deteriorate quickly due to exposure to adverse weather conditions, leading to increased maintenance costs.

Another critical thing to note is that surface pipes can move within openings when fluids flow through them. This is because the fluids’ forces can cause the movement of pipes. Similarly, thermal expansion and contraction are possible, especially in PVC pipes. Thus, openings for surface pipes can be opened up even when they are caulked.

725E2880 0190 4010 8E28 BDEED5A3D27C — Complex surface piping on a building

Conduit Piping

In construction, almost anything can be concealed in the structural components or members of buildings. So, pipes are included. Conduit piping involves concealing pipes in a building’s walls or other structural members. Unfortunately, conduit piping is mostly not considered during the structural design of building structures. Similarly, there are lots of conduit piping in buildings that occur as a result of afterthoughts.

IMG 20221019 WA00452 — *Pipes concealed in a slab and beam*

In most cases, conduit piping involves running pipes across the length of structural members or through them. Consequently, this reduces the load-carrying capacity of such members and the formation of honeycombs, especially around the pipes. Therefore, conduit piping has a higher probability of affecting the structural system, structural integrity, and building performance. However, with BIM and technical coordination, building services can be integrated during the design stage and most potential issues can be identified and resolved.

image 26fe7db0 9eae 4e44 81ef 153206d94f1820221019 062523 — *Building appearance adversely affected by leakage of concealed pipes*

Unlike surface piping, conduit piping can lead to significant demolitions, especially in cases of afterthoughts. Furthermore, drilling, cutting or breaking of walls or structural members induces vibrations that can affect such members’ structural stability and integrity. Lastly, it may be challenging to conduct maintenance works on concealed pipes. Thus, leakage is a significant concern because it can result in the corrosion of reinforcements and affect the durability of building components.

unsightly arrangement of MEP services — Unsightly conduit plumbing construction

Duct Piping

Duct piping is the new normal for modern building constructions. Duct piping is the best practice, and it solves all the challenges with surface and conduit piping methods. For duct piping, an architect working on a building design provides ducts, usually small spaces that run vertically through all the floors, to receive and house MEP services, including plumbing pipes. Thus, the probability of installing pipes due to afterthoughts is significantly reduced.

image 0adea431 cc2a 4e0d bcdf 8d94ea3303dd20221019 062525 — *Duct to house MEP services*

The advantages of duct piping cannot be overstated. It is a cost-effective method with less concern for leakage and maintenance. In addition, duct piping encourages durability, as plumbing pipes are shielded from adverse weather conditions, and it also improves the physical appearance of a building.

Best practices for duct piping

It is one thing to provide ducts in building structures, while it is another to do it right. The best practices for duct piping are listed below.

Group like services together. For instance, all water services can be together in a duct, while electrical and air-conditioning services can be put together in a duct.
A duct must be big enough to allow a person to move freely and work inside it, or good enough to allow someone to work from the exterior.
There must be a fixed metal ladder in each duct or provision for opening a duct at every floor level.
Ensure that ducts are rat-proof.
Ensure duct openings are from outside to prevent attending to service lines inside the building.
Keep all ducts locked for safety.

Conclusion

It is time we recognized that the conduit and surface piping methods had done more harm than good to buildings. Arguably, they have also increased the cost of maintenance of buildings. Conduit and surface piping have worked effectively in the past, but it seems that they no longer work in this age because a lot has changed in terms of the quality and durability of building materials.

The duct method of piping is the way forward. It is for our good and that of our clients if we eliminate the methods that affect building structures’ durability, performance, and structural integrity. With duct piping, we can be sure of reducing building maintenance costs and saving buildings from deterioration and eventual collapse.

Some Problems with Groundwater Lowering

By

Ubani Obinna

-

October 22, 2022

Table of Contents

Groundwater lowering operations may have an impact on water levels in a large area, even some distance away from the actual construction site. Although this usually doesn’t result in problems, there are some situations where unfavourable side effects such as subsidence could occur.

The following effects are covered in this article:

Settlement resulting from the instability of excavations when groundwater is not adequately controlled.
Ground settlements caused by loss of fines.
Ground settlements induced by increases in effective stress, and associated structural damage or distress.

Artificial groundwater recharge systems can be used to mitigate some of the negative consequences of groundwater lowering.

Settlement due to groundwater lowering

Every groundwater reduction operation will inevitably result in ground settlements. Most of the time, the settlements are so small that surrounding buildings show no distortion or damage. On occasion, however, settlements may be significant enough to cause structures to deform or stress in a detrimental way. This can range from modest architectural finish cracking to serious structural damage. In extreme circumstances, these effects have impacted numerous structures and have spread out hundreds of metres from the construction site itself.

A pre-construction building condition survey must be performed whenever there is any possibility that groundwater lowering (or any other construction operation) may cause ground settlements beneath existing structures. The objective of this activity often referred to as a dilapidation study, is to document the existing condition of any structures that may be impacted by settlement.

Groundwater decreasing may result in settlements for a variety of reasons, some of which are simple to avoid and some of which are more difficult to do so:

1 Settlement resulting from the instability of excavations when groundwater is not adequately controlled.
2 Settlement caused by loss of fines.
3 Settlement induced by increases in effective stress.

Settlement due to poorly controlled groundwater

Uncontrolled seepages, groundwater “blow,” and unstable excavations could result from inadequate groundwater lowering control. These issues could be the result of a number of factors, including a failure to recognise the importance of groundwater control, an improper attempt to cut costs by reducing or eliminating groundwater control from temporary works, a lack of standby or backup facilities to prevent pumping interruptions, and ground or groundwater conditions that were not taken into account during the site investigation or design or detected by construction monitoring.

Soil material will be washed into the excavation if there is a sudden “blow” or failure of the excavation. This has the potential to produce expansive settlements near the excavation that are unpredictable and far greater than the stress settlements linked to successful groundwater control. Any structures in the vicinity of the uncontrolled settlements are probably going to sustain significant damage.

Settlement due to loss of fines

A phenomenon known as “loss of fines” can cause settlement if a groundwater lowering system continuously pumps “fines” (particles the size of silt and sand) in the discharge water. In the early phases of pumping, most dewatering systems will pump fines as a more permeable zone forms around the well or sump.

However, if fines are pumped for an extended length of time, the removal of particles would loosen the soil and could result in the formation of subsurface erosion channels (sometimes referred to as “pipes”). Ground movements and settlement may be caused by compaction of the loosening soil or by the collapse of such erosion channels.

Continuous pumping of fines is not normally a problem with wellpoints, deep wells or ejectors, provided that adequate filter packs have been installed and monitored for fines in their discharge. Occasionally, a sand pumping well may be encountered, perhaps caused by a cracked screen or poor installation techniques. Such wells should be taken out of service immediately.

Sump pumping is the technique that results in fines being lost most frequently. This is due to the frequent neglect of installing sufficient filters surrounding sump pumps, which causes fine soil particles to become mobile as groundwater is pulled towards the pump.

Powers (1985) outlines the different types of soil where sump pumping should be avoided. These consist of:

Uniform fine sands
Soft non-cohesive silts and soft clays
Soft rocks where fissures can erode and enlarge due to high water velocities
Rocks where fissures are filled with silt, sand or soft clay, which may be eroded
Sandstone with uncemented layers that may be washed out.

Even the best-engineered sump pumping systems may experience issues with various types of soil. A method of groundwater lowering employing wells (wellpoints, deep wells, or ejectors) with properly designed and installed filters should be seriously considered.

Settlement due to increases in effective stress

As groundwater levels drop, pore water pressures naturally drop as well, increasing effective stress. The soil layer will compress as a result, resulting in ground settlements. The vast majority of the time, however, the effective stress settlements are so negligible that no harm is done to neighbouring structures.

The following variables will affect effective stress settlement:

The presence and thickness of a highly compressible layer of soil below the groundwater level, which will be affected by the pore water pressure reduction. Examples include soft alluvial silts and clays or peat deposits. The softer a soil layer (and the thicker it is), the greater the potential settlement.
The amount of drawdown. The greater the drawdown of the groundwater level, the greater the resulting settlement.
The period of pumping. In general, at a given site, the longer the pumping is continued, the greater the settlement.

Settlements caused by groundwater lowering will generally increase with time and will be greatest at the end of the period of pumping.

Light Gauge Steel Building Construction

By

Ubani Obinna

-

October 21, 2022

Table of Contents

Light gauge steel framing is typically based on the use of standard C or Z-shaped steel sections produced by cold rolling from strip steel. In general, hot-rolled steel sections used in fabricated steelwork, such as Universal Beams, are different from cold-formed steel sections. Galvanized steel with a typical thickness of 0.9 to 3.2 mm is used in cold-formed sections to prevent corrosion.

Cold-formed steel sections are widely utilised in many construction industries, including mezzanine floors, commercial, industrial, and hotel buildings. They are also becoming more popular in the residential market. In North America, Australia, and Japan, light steel frame is already well-established in the residential home construction industry. This article provides information on the various types of light gauge steel frame construction methods for residential buildings.

light gauge steel home under construction — **Figure 1:** Typical light gauge steel framing

Methods of Construction using Light Gauge Steel

Cold-formed sections, which can be prefabricated into panels or modules or constructed on-site using a variety of connecting techniques, are the fundamental building blocks of the light steel gauge frame. The various types of construction are discussed in the sections that follow.

‘Stick-build’ construction

In this construction approach (shown in Figure 2), discrete members are put together on-site to create columns, walls, rafters, beams, and bracing, which are then covered with cladding, internal lining, and other components. Although the elements are typically shipped pre-punched and cut to length, connections are done on-site using bolts, screws, or other suitable site procedures.

stick build light gauge construction method — **Figure 2:** Typical stick-build light gauge steel framing

The main advantages of ‘stick-build’ construction are:

construction tolerances and modifications can be accommodated on site
connection techniques are relatively simple
manufacturers do not require the workshop facilities associated with the panel or modular construction
large quantities of light steel members can be densely packed and transported in single loads
components can be easily handled on-site.

In comparison to the other methods, “stick-build” construction is typically labour-intensive on-site, but it can be effective in a complex building when prefabrication is difficult. In North America and Australia, where there is a strong infrastructure of trained contractors, this type of building is common. This is a result of the widespread use of power tools in the craft of timber frame construction. Traditional timber contractors have easily transitioned to light gauge steel frames in these nations.

Panel Construction

As shown in Figure 3, wall panels, floor cassettes, and roof trusses can be manufactured in a factory and then erected on-site. Panels are fabricated in specialised jigs for precision. To speed up building on-site, some finishing materials may be used in the factory. Steel pieces alone or with facing materials and insulation placed at the factory can make up panels. The panels are connected on-site utilising customary methods (bolts or self-drilling screws).

The main advantages of panel or sub-frame construction are:

speed of erection of the panels or sub-frames
quality control in production
reduced site labour costs
scope for automation in factory production.

The fact that the panels are prefabricated in a manufacturing environment improves their geometrical precision and dependability compared to stick-build construction. To achieve quick panel assembly and to reach the highest level of building efficiency, precise foundation planning and installation are essential.

Modular Construction

In modular construction, units can be brought to the site with all interior finishes, fixtures, and fittings already installed because they are totally constructed in the factory, as shown in Figure 4. To create a stable final construction, units may be piled one above the other or side by side.

Where massive production runs for the same basic configuration of modular unit are feasible, modular building is most cost-effective. This is possible because the costs of prototype and setup, which are largely scale-independent, may be distributed over numerous units.

Platform and ‘balloon’ construction

“Stick-build” or panel components can be put together in either “platform” or “balloon” construction. The walls are not physically continuous in platform construction since the floors and walls are created one level at a time. In some types of construction, loads are carried from the walls above to the walls below through the floor joists.

The wall panels used in “balloon” construction are frequently significantly larger and extend over multiple stories. These panels require temporary bracing while the floors are being placed since they are more challenging to assemble than single-storey height panels. The fundamental benefit of this strategy is that loads are carried directly from the walls above to those below. The external cladding or finishes are often installed and affixed to the frames on-site in both types of construction.

medium rise light gauge steel framing — **Figure 5:** Medium rise light gauge steel framing

The Difference Between Effective and Gross Section Properties

By

Ubani Obinna

-

October 21, 2022

In the design of steel structures, it is necessary to determine the section’s dimensional properties under consideration before a member’s resistance to bending, compression or other types of loading can be calculated. According to Eurocode 3, the section properties usually considered are the effective and gross section properties.

According to clause 6.2.2.1 of EN 1993-1-1, the properties of the gross cross-section should be determined using the nominal dimensions. Holes for fasteners need not be deducted, but allowance should be made for larger openings. Splice materials should not be included. When the effects of holes and openings are considered in the analysis, it is referred to as net section properties.

Effective section properties are the characteristics of a fictitious cross-section whose area has been reduced to account for the effects of local buckling. It can also be essential to make additional reductions to account for distortional buckling. The effective properties of the section are always used to compute the bending and compression resistances of light steel members (cold-formed steel sections such as that found in roof purlins).

light gauge steel sections — Light gauge steel sections (cold-formed sections)

Effective Section Properties

According to clause 6.2.2.4 of EN 1993-1-1, when the cross-sections with a class 3 web and class 1 or 2 flanges are classified as effective Class 2 cross-sections, see clause 5.5.2(1) of EN 1993-1-1, the proportion of the web in compression should be replaced by a part of 20εt_w adjacent to the compression flange, with another part of 20εt_w adjacent to the plastic neutral axis of the effective cross-section in accordance with Figure 6.3 of EN 1993-1-1 (shown below).

However, the effective cross-section properties of Class 4 cross-sections should be based on the effective widths of the compression parts. For cold-formed sections, it should be based on the requirements of EN 1993-1-3.

For light gauge steel sections, there is an implied presumption that the cross-section belongs to class 4 rather than classifying it (although this term is not used in BS EN 1993-1-3). Using the same methods as for class 4 sections to BS EN 1993-1-1, the design process concentrates on calculating the effective section properties after making this assumption. Effective section characteristics are used to reduce the amount of calculation necessary without too conserving on cross-section resistance by simplifying the intricate stress distributions related to local buckling.

The effective width approach substitutes simplified equivalent stresses acting over two equal widths of b_eff/2 for the actual stress distribution acting across element width b. It is assumed that there is no stress in the centre of the plate, which is the area most susceptible to local buckling, and it is completely disregarded. The end result is a straightforward model where uniform stress assumed to act over a narrower plate is equal to the steel’s yield strength.

Effective width concept — The effective width concept

The approach used by BS EN 1993-1-3 applies the above-illustrated effective width idea to the cross-section of a light gauge steel element. Each of the features of the cross-section—flanges, webs, lips, etc.—is treated as the flat plate in the Figure above. Each element that is under compressive stress has its effective width b_eff determined (either due to applied axial compression or bending).

The effective area of the element A_eff is then obtained by multiplying b_effby the section thickness t. Elements not subjected to compressive stress are not susceptible to local buckling, so the full element width b may be used in the calculation of the effective section properties.

Gross Section Properties

The term gross section properties, as the name implies, refers to the entire cross-section without any reduction for local buckling. For the majority of common section shapes, calculating the gross section properties only requires adding up the elemental areas and first and second moments of area (for flanges, web, stiffeners, etc.), determining the major and minor centroidal axes’ locations, and deriving the second moment of area for the entire section from these values. If necessary, a similar procedure can be repeated for additional properties.

Requirements for a Cofferdam

Planning a Cofferdam

Common Causes of Cofferdam Failures

Design of Cofferdams

Single skin Cofferdams

Walls/double skin Cofferdams

Cellular Cofferdams

Dispersive Soils in Construction

Sodicity and Dispersive Soils

References

Applicable Soil Types

The procedure of Stone Column Installation

Equipment

Materials

Design of Stone Columns

Quality assurance and quality control

Normalised Compressive Strength of Masonry

Compressive Strength of Mortar

Unit Manufacturing Control

Class of Execution Control

Characteristic Compressive Strength of Masonry

Partial Factor for Materials (γM)

Design of unreinforced masonry walls subjected to vertical loading

Effective Height of Masonry Wall

Worked Example on Masonry wall panel design

Unreinforced masonry walls subjected to lateral loading

Purpose of Digital Twin

Types of Digital Twin

Descriptive twin

Informative twin

Predictive twin

Comprehensive twin

Autonomous twin

Examples of digital twin

Benefits of digital twin

Conclusion

References

Design Example of Biaxial Eccentrically Loaded Pad Footing

Design Approach 1 (DA 1) – Combination 1

Bearing Resistance

Pad base pressure

Design Approach 1 (DA 1) – Combination 2

Bearing resistance (Section 6.5.2)

Pad base pressure

Foundation design (EN1992-1-1:2004)

Concrete details

Reinforcement details

Rectangular section in flexure (x-axis)

Rectangular section in shear (x-axis)

Rectangular section in flexure (y-axis)

Crack control

Rectangular section in shear (y-axis)

Punching shear

Punching shear perimeter at column face

Punching shear perimeter at 1d from column face

Punching shear perimeter at 2d from column face

Piping Methods

Surface Piping

Conduit Piping

Duct Piping

Best practices for duct piping

Conclusion

Settlement due to groundwater lowering

Settlement due to poorly controlled groundwater

Settlement due to loss of fines

Settlement due to increases in effective stress

Methods of Construction using Light Gauge Steel

‘Stick-build’ construction

Panel Construction

Modular Construction

Platform and ‘balloon’ construction

Effective Section Properties

Gross Section Properties

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Partial Factor for Materials (γ_M)