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Structural Analysis and Design of Truss Bridges

Lattice truss structural systems have been employed in constructing railway and highway bridges with great success for so many years. The design of truss bridges involves the analysis of the structure to obtain the internal forces due to moving traffic and permanent loads (self-weight), selection of adequate steel members, design of the connections, and check for fatigue. The availability of numerous commercial design software has made the analysis and design of 3D truss bridges easier than it was in the past.

The Warren truss, the Modified Warren truss, and the Pratt truss are the three major truss configurations in use today, and they can all be employed as an underslung truss, a semi-through truss, or a through truss bridge.

TRUSS BRIDGE 1

In an underslung truss, the live loading caused by the passage of automobiles or trains is carried directly by the top chord. In situations where the depth of construction or clearance under the bridge is not critical, underslung trusses can be conveniently used.

In semi-through trusses, vehicles travel on the bottom chord of the truss, but the transient live load projects above the top chord members due to the height of the vehicles relative to the top chord of the truss. As a result, the top chords in semi-through trusses cannot be braced laterally, and these chord elements must rely on U-frame action for lateral stability. However in a through truss bridge, vehicles travel through the centre of the bridge on the bottom chord, and the space between the live load and the top chords is sufficient that the top chord members can be braced laterally. Through truss bridge appears to be the most common type of truss bridge.

Types of truss bridges
Figure 1: Major types of truss bridges (Parke and Harding, 2008)

Internal Forces in Truss Bridges

The members of a truss bridge will predominantly carry axial tension or axial compression stresses if the structure is designed and detailed so that live loading is effectively applied at the nodes. The global bending moment acting on the bridge may be resolved into a couple made up of the compression forces in the top chord and axial tension forces in the bottom chord. Similarly, the diagonal web elements carry the global shear force exerted on the truss bridge, either in axial tension or compression, depending on the configuration of the truss.

UNDERSLUNG TRUSS BRIDGE

As an example, the diagonal web elements of a Warren truss alternately carry compression and tension over the bridge. The internal diagonal web members of a Pratt truss, on the other hand, are all loaded in tension, while the shorter vertical web members are loaded in compression.

Members of Truss Bridges

The chords and web members of truss bridges can be made out of a variety of steel sections. For the tension and compression chords as well as the web members of short-span  (30–50 m) highway trusses, rolled “H” sections and square hollow sections are suitable. Larger fabricated sections, such as a “top hat” section or box section, will be needed for the chords of longer highway truss bridges or trusses bearing railway loads.

Built-up through truss bridge
Built-up through truss bridge

Analysis of Truss Bridges

Truss bridges transmit imposed loads to the foundations through the axial tension and compression forces in the members. Therefore, these structures can be analyzed as pin-jointed members, either as a two-dimensional truss or, more preferably, as a three-dimensional space truss.

This form of analysis assumes that member connections are pinned, which means that none of the truss members may attract moment or torsion. By hand, a two-dimensional plane truss analysis can be solved by utilizing equilibrium equations to resolve the forces at each joint in turn, or by employing the method of sections to free-body segments of the bridge truss, again using equilibrium equations to derive member forces.

The stiffness method can also be used to calculate node displacements first, and then member forces. Nowadays, truss bridges do not have pinned joints; instead, the connections are welded or bolted; yet, analyzing the structure as a two- or three-dimensional pin-jointed assembly allows for an accurate assessment of member axial stresses but overpredicts truss node displacements.

PROFILE OF THROUGH TRUSS BRIDGE

However, since the joints are not pinned in real-life construction, it is necessary to analyze the truss as a three-dimensional space frame with six degrees of freedom at each node in order to obtain a more realistic prediction of node displacements as well as an assessment of the secondary bending and torsion moments, which will be small but still present.

Secondary moments and torsions acting on the structure can affect the bridge’s fatigue life, particularly if the truss is continuous and spans multiple supports. By guaranteeing that the neutral axes of all members meeting at a node intersect at a single location in three-dimensional space, secondary forces and hence stresses can be reduced.

Worked Example: Design of Truss Bridge

A 9.0m wide through-truss bridge is to be designed to carry normal traffic across a river. The total height of the bridge is 5m, and I-sections are to be utilised in the top and bottom chord members of the truss, while square hollow sections will be utilised for the web members. The vertical members of the web are spaced at 2.5m each and the total length of the bridge is 25m.

image 16
Modified Warren Truss Bridge

The deck of the bridge is composed of primary and secondary steel beam members. The floor beams consist of UB 457x191x161 members supported by the UB 610x305x179 bottom chord rail of the trusses and spaced at 2.5m intervals. The stringers are the secondary UB 305x102x28 members running parallel to the bottom chord and spaced at 1.5m. A 200mm thick reinforced concrete deck is expected to sit on the beams.

TRUSS BRIDGE DECK

The truss bridge has been modelled on Staad Pro software as shown below. The top of the truss bridge (top chord) will be braced using UB 254x146x37 members in a K-truss arrangement (see below) to restrain the top chord from sway under wind action.

Structural Model of a Truss Bridge
Structural Model of a Truss Bridge

Loading

In this article, the truss bridge will be analysed for the self-weight (all

dead and superimposed loads) and traffic load. All other environmental loads and indirect actions will not be considered.

Dead Load
(1) Self-weight of steel members (to be calculated automatically by Staad Pro)
(2) Self-weight of 200mm thick reinforced concrete deck = 0.2 × 25 = 5 kN/m2
(3) Self-weight of 75mm thick asphalt wearing course = 0.075 × 23.5 = 1.8 kN/m2

Total pressure dead load = 6.8 kN/m2

Live Load
According to the requirements of Load Model 1 (LM 1), the carriageway width of 9m can be divided into three notional lanes as shown below;

load model 1 on truss bridge

In essence, the traffic on the bridge will be represented by the UDL as specified above and the tandem system. The worst effect of the wheel load on the bridge deck will be considered. However, it is important that the influence line analysis of the bridge be carried out, in order to determine the wheel load location that will produce the worst effects on the structure.

image 5

Bottom Chord Analysis Results

The result below depicts the internal stresses induced in the bottom chord at the load combination 1.35gk + 1.5qk (where qk represents the UDL component of the traffic load only).

image 8

The result below depicts the internal stresses induced on the bottom chord under the unfactored moving tandem wheel load only.

image 9

A little consideration will show that the following results are applicable for the bottom chord;

Design Axial compression (member 1): 462.892 + 1.5(300.842) = 914.155 kN
Design axial tension (members 5 and 6) = 457.486 + 1.5(321.591) = 939.8725 kN
Design major bending moment (member 3): 231.737 + 1.5(231.617) = 579.1625 kNm
Design minor axis bending moment = 2.541 + 1.5(1.590) = 4.926 kNm
Design Shear (major axis) = 101.619 + 1.5(158.288) = 339.051 kN
Design Shear (minor axis) = 1.853 + 1.5(1.17) = 3.609 kN

It is obvious that these maximum forces do not interact on the same point in the section. However, for the sake of simplicity, let us assume they interact at the same point in the structure. The design verifications are as follows;

Design of the Bottom Chord

Section s1 results summaryUnitCapacityMaximumUtilisationResult
Shear resistance (y-y)kN1441.9339.10.235PASS
Shear resistance (z-z)kN2047.73.60.002PASS
Bending resistance (y-y)kNm1470.0579.20.394PASS
Bending resistance (z-z)kNm303.14.90.016PASS
Compression resistancekN5598.0914.20.163PASS
Comb. bending and axial force   0.571PASS

Section details
Section type; UB 610x305x179 (BS4-1)
Steel grade – EN 10025-2:2004;  S275
Nominal thickness of element;  tnom = max(tf, tw) = 23.6 mm
Nominal yield strength;  fy = 265 N/mm2
Nominal ultimate tensile strength; fu = 410 N/mm2
Modulus of elasticity; E = 210000 N/mm2

Classification of cross sections – Section 5.5
ε = √[235 N/mm2 / fy] = 0.94

Internal compression parts subject to bending and compression – Table 5.2 (sheet 1 of 3)
Width of section; c = d = 540 mm
α = min([h / 2 + NEd / (2 × tw × fy) – (tf + r)] / c, 1) = 0.727
c / tw = 38.3 = 40.7ε ≤ 396ε / (13α – 1); Class 1

Outstand flanges – Table 5.2 (sheet 2 of 3)
Width of section; c = (b – tw – 2r) / 2 = 130 mm
c / tf = 5.5 = 5.8ε ≤ 9ε; Class 1
Section is class 1

Check compression – Section 6.2.4
Design compression force; NEd = 914.2 kN
Design resistance of section – eq 6.10;                        
Nc,Rd = Npl,Rd = Afy / γM0 = 6044.2 kN
NEd / Nc,Rd = 0.151

PASS – Design compression resistance exceeds design compression

Slenderness ratio for y-y axis flexural buckling – Section 6.3.1.3
Critical buckling length; Lcr,y = Ly_s1 = 2500 mm
Critical buckling force;  Ncr,y = π2EIy / Lcr,y2 = 507456.5 kN
Slenderness ratio for buckling – eq 6.50; λy = √(Afy / Ncr,y) = 0.109

Check y-y axis flexural buckling resistance – Section 6.3.1.1
Buckling curve – Table 6.2; a
Imperfection factor – Table 6.1;αy = 0.21
Buckling reduction determination factor; φy = 0.5[1 + αyy – 0.2) + λy2] = 0.496
Buckling reduction factor – eq 6.49; χy = min[1 / (φy + √(φy2 – λy2)), 1] = 1

Design buckling resistance – eq 6.47;                          
Nb,y,Rd = χyAfy / γM1 = 6044.2 kN
NEd / Nb,y,Rd = 0.151

PASS – Design buckling resistance exceeds design compression

Slenderness ratio for z-z axis flexural buckling – Section 6.3.1.3
Critical buckling length; Lcr,z = Lz_s1 = 2500 mm
Critical buckling force; Ncr,z = π2EIz / Lcr,z2 = 37832.1 kN
Slenderness ratio for buckling – eq 6.50;  λz = √(Afy / Ncr,z) = 0.4

Check z-z axis flexural buckling resistance – Section 6.3.1.1
Buckling curve – Table 6.2; b
Imperfection factor – Table 6.1; αz = 0.34
Buckling reduction determination factor;
φz = 0.5(1 + αzz – 0.2) + λz2) = 0.614

Buckling reduction factor – eq 6.49;
χz = min(1 / (φz + √(φz2 – λz2)), 1) = 0.926

Design buckling resistance – eq 6.47;                          
Nb,z,Rd = χzAfy / γM1 = 5598 kN
NEd / Nb,z,Rd = 0.163

PASS – Design buckling resistance exceeds design compression

Check of torsional and torsional-flexural buckling showed that the section is okay. – Section 6.3.1.4

Check for shear – Section 6.2.6
Height of web; hw = h – 2tf = 573 mm;                  
η = 1.000
hw / tw = 40.6 = 43.2ε/η < 72ε/η

Shear buckling resistance can be ignored

Design shear force;
Vy,Ed = 339.1 kN

Shear area – cl 6.2.6(3);
Av = max(A – 2btf + (tw + 2r)tf, ηhwtw) = 9425 mm2

Design shear resistance – cl 6.2.6(2);                          
Vc,y,Rd = Vpl,y,Rd = Av(fy /√3) / γM0 = 1441.9 kN
Vy,Ed / Vc,y,Rd = 0.235

PASS – Design shear resistance exceeds design shear force

Check bending moment – Section 6.2.5
Design bending moment; My,Ed = 579.2 kNm
Design bending resistance moment – eq 6.13;           
Mc,y,Rd = Mpl,y,Rd = Wpl.yfy / γM0 = 1470 kNm
My,Ed / Mc,y,Rd = 0.394

PASS – Design bending resistance moment exceeds design bending moment

Check bending and axial force – Section 6.2.9
Bending and axial force check – eq.6.33 & eq.6.34;  
Ny,lim = min(0.25Npl,Rd, 0.5hwtwfy / γM0) = 1070.5 kN
NEd / Ny,lim = 0.854

Allowance need not be made for the effect of the axial force on the plastic resistance moment about the y-y axis

Bending and axial force check – eq.6.35; 
Nz,lim = hwtwfy / γM0 = 2141.0 kN
NEd / Nz,lim = 0.427

Allowance need not be made for the effect of the axial force on the plastic resistance moment about the z-z axis
αN = 2
βN = max(5n, 1) = 1

For bi-axial bending – eq.6.41;                                      
[My,Ed / Mpl,y,Rd]αN + [Mz,Ed / Mpl,z,Rd]βN = 0.171

PASS – Biaxial bending utilisation is acceptable

Check combined bending and compression – Section 6.3.3
Equivalent uniform moment factors – Table B.3;        
Cmy = 1.000
Cmz = 1.000
CmLT = 1.000

Interaction factors kij for members susceptible to torsional deformations – Table B.2
Characteristic moment resistance; 
My,Rk = Wpl.yfy = 1470 kNm

Characteristic moment resistance;                              
Mz,Rk = Wpl.zfy = 303.1 kNm

Characteristic resistance to normal force;                  
NRk = Afy = 6044.2 kN

Interaction factors;                                                           
kyy = Cmy(1 + min(λy – 0.2, 0.8) × NEd / (χyNRk / γM1)) = 0.986
kzy = min(0.6 + λz, 1 – 0.1λzNEd / ((CmLT – 0.25) × χzNRk / γM1)) = 0.991
kzz = Cmz(1 + min(2λz – 0.6, 1.4) × NEd / (χzNRk / γM1)) = 1.033
kyz = 0.6kzz = 0.620

Interaction formulae – eq 6.61 & eq 6.62;                    
NEd / (χyNRk / γM1) + kyyMy,Ed / (cLTMy,Rk / γM1) + kyzMz,Ed / (Mz,Rk / γM1) = 0.55
NEd / (χzNRk / γM1) + kzyMy,Ed / (cLTMy,Rk / γM1) + kzzMz,Ed / (Mz,Rk / γM1) = 0.571

PASS – Combined bending and compression checks are satisfied.

Design of the Web Members (Verticals and Diagonals)

The result below depicts the internal stresses induced in the bottom chord at the load combination 1.35gk + 1.5qk (where qk represents the UDL component of the traffic load only).

image 13

The result below depicts the internal stresses induced on the bottom chord under the unfactored moving tandem wheel load only.

image 14

Design Axial compression : 1181.948 + 1.5(761.358) = 2324 kN
Design axial tension = 928.940 + 1.5(623.655) = 1864 kN
Design major bending moment: (ignored for brevity)
Design minor axis bending moment = (ignored for brevity)
Design Shear (major axis) = (ignored for brevity)
Design Shear (minor axis) = (ignored for brevity)

Classification of cross sections – Section 5.5
ε = √[235 N/mm2 / fy] = 0.92

Internal compression parts subject to compression – Table 5.2 (sheet 1 of 3)
Width of section;                                                              
c = b – 3t = 212.5 mm
c / t = 17 = 18.4ε <= 33ε; Class 1

Internal compression parts subject to compression – Table 5.2 (sheet 1 of 3)
Width of section;                                                              
c = h – 3t = 212.5 mm
c / t = 17 = 18.4ε <= 33ε; Class 1
Section is class 1

Check compression – Section 6.2.4
Design compression force; NEd = 2324 kN
Design resistance of section – eq 6.10;
Nc,Rd = Npl,Rd = Afy / γM0 = 3219.5 kN
NEd / Nc,Rd = 0.722

PASS – Design compression resistance exceeds design compression

Slenderness ratio for y-y axis flexural buckling – Section 6.3.1.3
Critical buckling length; Lcr,y = Ly_s1 = 5590 mm (considering the length of the diagonal web members)

Critical buckling force;                                                    
Ncr,y = π2EIy / Lcr,y2 = 7239.9 kN

Slenderness ratio for buckling – eq 6.50;                     
λy = √(Afy / Ncr,y) = 0.667

Check y-y axis flexural buckling resistance – Section 6.3.1.1
Buckling curve – Table 6.2; a
Imperfection factor – Table 6.1; ay = 0.21

Buckling reduction determination factor;
φy = 0.5[1 + αyy – 0.2) + λy2] = 0.771

Buckling reduction factor – eq 6.49;                              
χy = min[1 / (φy + √(φy2 – λy2)), 1]= 0.863

Design buckling resistance – eq 6.47;                          
Nb,y,Rd = χyAfy / γM1 = 2777.7 kN
NEd / Nb,y,Rd = 0.837

PASS – Design buckling resistance exceeds design compression

For completeness, the section should be checked for shear, torsional buckling, and axial/moment interaction.

Conclusion

The method adopted in this article is suitable for draft/preliminary designs. The approach can be extended and used to design and select all the members of the truss bridge. For instance, a little review of the design of the bottom chord shows that there is still room for reduction of the member size, while the same cannot be said for the web members.

The floor beams of the section will need to be designed as a composite beam, taking into account the interaction of the concrete deck. After all the members have been selected and checked, a detailed/final analysis and design can be carried out, to verify the suitability of the selected members.



Design of Composite Beams (AISC 360-16)

The general theory of composite beam design calculations is well known among structural engineers, however, the execution of composite beam design in practice necessitates taking into account a number of factors in addition to structural calculations, such as fire engineering, constructability, and more. This article discusses the structural design of composite beams and some of the factors that must be taken into account while designing composite beams.

The construction industry in the United States of America now uses two basic approaches to composite beam design – The LRFD and ASD methods. The method featured in the 3rd Edition LRFD Manual of Steel Construction is both simpler in design and more cost-effective than the method described in the 9th Edition Manual of Steel Construction (ASD).

In the ASD method, the moment capacity is computed from the superposition of elastic stresses, while in the LRFD approach, the moment capacity is computed from the distribution of plastic stresses.

composite construction

It is usually possible to produce an economical design with partial composite action in the beam. In many design situations, increasing the beam size can satisfy the design moment while significantly reducing the number of studs needed. The design of composite beams is almost always carried out using computer software or design tools like those found in the AISC Manual.

The deck size should, within reason, be chosen to allow for the beam spacing. For un-shored construction, the Steel Deck Institute (SDI) offers tables that show the maximum span permitted for a specific deck and slab arrangement. In general, the economy of the steel floor system is improved by maximizing the span for a given deck size, for un-shored construction. It is advised that you choose a deck assuming a 2-span un-shored condition and avoid single-span situations as much as possible.

The ponding of concrete as well as how the slab is poured must be taken into account. You might want to factor in an extra 1/2 inch of concrete to accommodate for ponding when estimating the amount of concrete needed to construct level slabs. Since the wet weight of lightweight concrete has been reported in the field to range up to 125 pcf, it is crucial to consider this.

Materials for Composite Beam Construction

All of the approved ASTM material specifications for the construction of composite floors are included in Section A3 of the AISC Specification. During the design of composite beams, it is pertinent to always specify ASTM A992 when broad flange beams are being used. However, HSS, pipes, and built-up shapes are also covered by the AISC requirements. The ASTM A108 shear stud, which has a tensile strength of 60 ksi, is frequently used in specifications. 3/4-inch diameter studs are the most typical size used in building construction.

Composite beam construction

In addition to reinforcing bars and welded wire, the composite slab may also be steel fibre reinforced in accordance with ASTM C1116 in specific circumstances. For normal-weight concrete and light-weight concrete, the minimum specified compressive strength of the concrete in the slab must be between 3 ksi and 10 ksi. Higher strengths should only be relied upon for rigidity. In order to comply with standard fire-rated assemblies, 3.5 ksi normal-weight concrete and 3 ksi light-weight concrete are typically specified.

Cambering in Composite Beams

Although there are several ways to obtain level steel-framed floors, cambering of beams is the technique of choice in the United States. Engineers in the field frequently misunderstand the purposes of proper beam cambering. Beam camber is just one component of a comprehensive floor levelness strategy that must take into account the slab pour method, building occupancy, and steel fabrication and installation procedure.

The main objective of cambering beams is to accurately predict how much the beam will actually deflect under the weight of the concrete. Correct camber is best attained between 75 and 80 percent of the estimated dead load deflection because of connection restraints and fabrication tolerances. Beams should never have excessive camber. Additionally, cambering is improper for a variety of beam types, including brace beams and very short beams.

Serviceability of Composite Beams

For composite floors, serviceability factors to be taken into account are long-term deflections from the superimposed dead load, short-term deflections from the live load, vibration control, and slab system performance. The acceptance standards relevant to the intended floor use, creep deflections under superimposed dead load, and partial composite action must all be taken into account when evaluating deflections.

Design and Detailing of Studs

The 1999 AISC Specification’s Section 15.6 addresses proper stud design and detailing. In the longitudinal and transverse directions, the minimum stud spacing is 6 times and 4 times the stud diameter, respectively. Two new factors—stud geometry and stud position inside the deck ribs—will need to be taken into account in accordance with the 2005 Specification.

image 1

Design Example of Composite Beams

Design a 25ft long secondary beam in a proposed commercial complex. The deck ribs are perpendicular to the beam, and the secondary beams are spaced at 10 ft intervals. The concrete for the steel deck has an overall depth of 6 inches with a compressive strength of 3 ksi. The following loadings are anticipated on the floor;

Weight of steel deck = 3.000 psf
Additional dead load= 20.000 psf
Weight of steel beam = 54.000 lb/ft
Weight of construction live load = 20.000 psf
Floor live load = 40.000 psf
Lightweight partition load = 10.000 psf

composite structure

Basic dimensions
Beam span;  L = 25.000 ft
Beam spacing on one side; b1 = 10.000 ft
Beam spacing on the other side;  b2 = 10.000 ft
Deck orientation; Deck ribs perpendicular to beam

image 2

Profiles are assumed to meet all dimensional criteria in AISC 360-16

Overall depth of slab;   t = 6.000 in
Height of ribs; hr = 1.500 in
Centers of ribs; ribccs = 6.000 in
Average width of rib; wr = 2.500 in

Material properties

Concrete
Specified compressive strength of concrete;  f’c = 3.00 ksi
Wet density of concrete; wcw = 150 lb/ft3
Dry density of concrete;  wcd = 130 lb/ft3                                    
Modulus of elasticity of concrete; Ec = wcd1.5 × √(f’c × 1 ksi) /(1 lb/ft3)1.5 = 2567 ksi

Steel
Specified minimum yield stress of steel; Fy = 50 ksi
Modulus of elasticity of steel; ES = 29000 ksi

Loading – secondary beam

Weight of slab construction stage; wslab_constr = [t – hr × (1 – wr / ribccs)] × wcw  = 64.062 psf
Weight of slab composite stage; wslab_comp = [t – hr × (1 – wr / ribccs)] × wcd  = 55.521 psf
Weight of steel deck; wdeck = 3.000 psf
Additional dead load;  wd_add = 20.000 psf
Weight of steel beam;  wbeam_s = 54.000 lb/ft
Weight of construction live load;  wconstr = 20.000 psf
Floor live load; wimp = 40.000 psf
Lightweight partition load; wpart = 10.000 psf

Total construction stage dead load;                             
wconstr_D = [(wslab_constr + wdeck + wd_add) × ((b1+b2)/2)] + wbeam_s = 924.625 lb/ft

Total construction stage live load;                                
wconstr_L = wconstr × (b1 + b2) / 2 = 200.000 lb/ft

Total composite stage dead load(excluding walls);  
wcomp_D = [(wslab_comp + wdeck + wd_add + wserv) × (b1 + b2)/2] + wbeam_s = 839.208 lb/ft

Total composite stage live load;                                   
wcomp_L = (wimp + wpart) × (b1 + b2)/2 = 500.000 lb/ft;

Design forces – secondary beam

Max ultimate moment at construction stage;              
Mconstr_u = ( 1.2wconstr_D + 1.6wconstr_L ) × L2/ 8 = 111.684 kips_ft

Max ultimate shear at the construction stage;                   
Vconstr_u = ( 1.2wconstr_D + 1.6wconstr_L ) × L/2 = 17.869 kips

Maximum ultimate moment at the composite stage;
Mcomp_u = ( 1.2wcomp_D + 1.6wcomp_L ) × L2/ 8 + 1.2 × ww_par × L2/8 + 1.2ww_perp × (b1 + b2)/2 × L/4 = 141.176 kips_ft

Maximum ultimate shear at the composite stage;
Vcomp_u = ( 1.2wcomp_D + 1.6wcomp_L ) × L/2 + 1.2 × ww_par × L/2 + 1.2ww_perp × (b1 + b2)/2 × 1/2= 22.588 kips

Point of max. B.M. from nearest support;                    
LBM_near =  L/2 = 12.50 ft

Steel section check

Trial steel section; W10X54
Plastic modulus of steel section; Zx = 66.60 in3
Elastic modulus of steel section; Sx = 60.00 in3
Width to thickness ratio; λf = bf / ( 2tf ) = 8.130
Limiting width to thickness ratio (compact); λpf = 0.38 × √(ES / Fy) = 9.152
Limiting width to thickness ratio (non-compact); λrf = √(ES / Fy) = 24.083
Flange is compact

λw = h_to_tw = 21.200
Depth to thickness ratio (h/tw);  λw = 21.200
Limiting depth to thickness ratio (compact); λpw= 3.76 × √(ES / Fy) = 90.553
Limiting depth to thickness ratio (noncompact); λrw= 5.70 × √(ES / Fy) = 137.274
Web is compact

Strength check at the construction stage for flexure

Check for flexure

Plastic moment for steel section; Mp = FyZx = 277.500 kip_ft
Resistance factor for flexure; φb = 0.90

Design flexural strength of steel section alone;         
Mconstr_n = fb × Mp = 249.750 kip_ft

Required flexural strength;  Mconstr_u = 111.684 kip_ft
PASS – Beam bending at construction stage loading

Strength check at the construction stage for shear

Web area; Aw = d × tw = 3.737 in2
Web plate buckling coefficient;  kv = 5.34
Depth to thickness ratio (h/tw); λw = 21.200
Web shear coefficient; Cv1 = 1.00
Resistant factor for shear;  φv = 1.0
Design shear strength; Vconstr_n = φv × (0.6Fy × Aw × Cv1) = 112.110 kips
Required shear strength; Vconstr_u = 17.869 kips
PASS – Beam shear at construction stage loading

Design of shear connectors

Note – for non-uniform stud layouts a higher concentration of studs should be located towards the ends of the beam

Effective slab width of composite section;                  
b = min(L/8, b1/2) + min(L/8, b2/2) = 75.000 in
Effective area of concrete flange; Ac = b(t – hr) = 337.50 in2
Diameter of shear stud; dia = 0.750 in
Length of shear stud after weld; Hs = 3.00 in
Specified tensile strength of shear stud; Fu = 65 ksi
Cross section area of one shear stud; Asc = π × dia2 / 4 = 0.442 in2
Maximum diameter permitted;  diamax = 2.5 × tf = 1.537 in

PASS – Diameter of shear stud provided is OK

Point of max. B.M. from nearest support; 
LBM_near = 12.50 ft

No. of ribs from points of zero to max moment;         
ribnumbers = int(LBM_near /ribccs -1) = 24
No. of ribs with 1 stud per rib; Nr1 = 24
No. of ribs with 2 studs per rib; Nr2 = 0
No. of ribs with 3 studs per rib; Nr3 = 0

Total number of studs; Nprov = Nr1 + 2Nr2 + 3Nr3 = 24

Group effect factor for 1 stud per rib; Rg1 = 1.00
Group effect factor for 2 studs per rib; Rg2 = 0.85
Group effect factor for 3 studs per rib; Rg3 = 0.70

Value of emid-ht is less than 2 in (51 mm)

Position effect factor for deck perpendicular; Rp = 0.60

Nom. strength of one stud with 1 stud per rib;            
Qn1 = min(0.5 × Asc × √(f’c × Ec) , Rg1 × Rp × Asc × Fu ) = 17.230 kips

Nom. strength of one stud with 2 studs per rib;          
Qn2 = min(0.5 × Asc × √(f’c × Ec) , Rg2 × Rp × Asc × Fu ) = 14.645 kips

Nom. strength of one stud with 3 studs per rib;          
Qn3 = min(0.5 × Asc × √(f’c × Ec) , Rg3 × Rp × Asc × Fu ) = 12.061 kips

Total strength of provided shear connectors;             
Ssc = Nr1Qn1 + 2Nr2Qn2 + 3Nr3Qn3 = 413.51 kips

Resistance of concrete flange; Ccf = 0.85f’cAc = 860.625 kips
Resistance of steel beam; Tsb = AFy = 790.000 kips
Beam/slab interface shear force; C = min(Ccf, Tsb) = 790.000 kips

The strength of studs is less than the maximum interface shear force therefore partial composite action takes place

Strength check at partial composite action

Actual net tensile force; Vh = C = 790.000 kips

Assuming a plastic neutral axis (PNA) at the bottom of the steel beam flange.

Resultant compressive force at flange bottom;          
Pyf = bf × tf × Fy = 307.500 kips

Net force at steel and concrete interface;                   
Cnet = Tsb – 2Pyf = 175.000 kips

PNA is in the flange of I Section

Shear connection force;                                                 
Fshear = Ssc = 413.51 kips

Total depth of concrete at full stress;                           
dc = Fshear / (0.85 × f’c × b) = 2.162 in

Depth of compression from top of the steel flange;   
t’ = A / (2 × bf ) – 0.85f’c / Fybdc / (2 × bf ) = 0.376 in

Tension
Bottom flange component;                                            
Fbf = Fybf × tf = 307.500 kips

Moment capacity of bottom flange;                              
Mbf = Fbf(d – (tf /2) – t’) = 241.285 kip_ft

Web component;                                                             
Fweb = Fy(A – (2bf × tf ))= 175.000 kips

Moment capacity of web;                                               
Mweb = Fweb[((d – 2tf)/2)+ tf – t’] = 68.155 kip_ft

Top flange component;                                                  
Ftf_t = Fybf × (tf – t’) = 119.256 kips

Moment capacity of top flange;                                     
Mtf_t = Ftf_t (tf – t’)/2 = 1.185 kip_ft

Compression
Top flange component;                                                  
Ftf_c = Fybf × t’ = 188.244 kips

Moment capacity of top flange;                                     
Mtf_c = Ftf_ct’/2 = 2.953 kip_ft

Concrete flange component;                                         
Fcf = 0.85f’c × bdc = 413.512 kips

Moment capacity of concrete flange;                           
Mcf = Fcf(t – dc/2 + t’) = 182.476 kip_ft

Design flexural strength of beam;                                
Mcomp_n = fb( Mbf + Mweb + Mtf_t + Mtf_c + Mcf) = 446.450 kip_ft

Required flexural strength;                                            
Mcomp_u = 141.176 kip_ft

PASS – Beam bending at partial composite stage

Check for shear
Design shear strength;                                                   
Vcomp_n = Vconstr_n = 112.110 kips

Required shear strength;                                                
Vcomp_u = 22.588 kips

PASS – Beam shear at partial composite stage loading

Check for deflection (Commentary section 13.1)

Calculation of immediate construction stage deflection;

Deflection due to dead load;                                          
Dshort_D = 5 × wconstr_D × L4 / (384 × ES × Ix) = 0.9248 in

Amount of beam camber; Dcamber = 0.000 in

PASS – The camber is less than the construction stage dead load deflection

Deflection due to construction live load;                      
D2 = 5 × wconstr_L × L4 / (384 × ES × Ix) = 0.2000 in

Net total construction stage deflection;                       
Dshort = Dshort_D + D2 – Dcamber = 1.125 in

For short-term loading:-

Short-term modular ratio;                                               
ns = ES / Ec = 11.3

Depth of neutral axis from the top of concrete;                 
ys = [b(t – hr)/ns(t – hr)/2 + A(t + d/2)] / [b(t – hr)/ns + A]
ys = 5.294 in

Moment of inertia of fully composite section;
Is = Ix + A(d/2 + t – ys)2 + b(t – hr)3/(12ns) + b(t – hr)/ns(ys – (t – hr)/2)2
Is = 1154 in4

Fshear = Ssc = 413.5 kips

Effective of inertia for partially composite;            
Is_eff = 0.75[Ix + √(Fshear / C) × (Is – Ix)] = 688.9 in4

Proportion of live load which is short term; rL_s = 67 %

Deflection due to short-term live load;                         
DL_s = 5rL_swcomp_LL4 / (384ESIs_eff) = 0.1474 in

For long-term loading:

Long term concrete modulus as % of short term; rE_l = 50 %

Long-term modular ratio;                                                
nl = ES / (EcrE_l) = 22.6

Depth of neutral axis from top of concrete;                 

yl = [b(t – hr)/nl × (t – hr)/2 + A(t + d/2)] / [b(t – hr)/nl + A]
yl = 6.773 in

Moment of inertia of fully composite section;
Il = Ix + A(d/2 + t – yl)2 + b(t – hr)3/(12nl) + b(t – hr)/nl (yl – (t – hr)/2)2
Il = 923 in4

Effective moment of inertia for partially composite;            
Il_eff = 0.75[Ix + √(Fshear / C)(Il – Ix)] = 563.6 in4

Proportion of live load which is long term;                   
rL_l = 1 – rL_s = 33 %

Deflection due to long-term live load;                           
DL_l = 5 × rL_l ´ wcomp_L × L4 / (384 × ES × Il_eff) = 0.0887 in

Dead load due to parallel wall & superimp. dead;     
wD_part = ww_par + (wserv(b1+ b2) / 2) = 0.0000 lb/ft

Long-term deflection due to superimposed dead load (after concrete has cured)
Wall parallel to span and superimposed dead;          
D4 =5 × (wD_part) × L4 / (384 × ES × Il_eff) = 0.0000 in

Wall perpendicular to span;                                           
D5 =(ww_perp(b1+ b2) / 2) × L3 / (48 × ES × Il_eff) = 0.0000 in

Combined deflections
Net total construction stage deflection;                       
Dshort = Dshort_D + D2 – Dcamber = 1.125 in

Net total long-term deflection;                                       
Dlong = Dshort_D + DL_s + DL_l + D4 + D5 – Dcamber = 1.161 in

Combined short and long-term live load deflection;     
Dlive = DL_s + DL_l = 0.236 in

Net long-term dead and superimposed dead deflection; 
Ddead = Dshort_D +D4 + D5 – Dcamber = 0.925 in

Post composite deflection;                                            
Dcomp = DL_s + DL_l + D4 + D5 = 0.236 in

Allowable max deflection; 
DAllow = 1.250 in

PASS – Deflection less than allowable

Sanitary Landfills

Household wastes are usually dumped in municipal solid waste landfills (MSWLFs). Landfills are sites that are designed for the dumping and management of municipal solid wastes. However, non-hazardous sludge, industrial solid waste, and construction and demolition waste can be dumped in landfills as well.

Modern landfills are well-engineered structures that are situated, developed, managed, and monitored to ensure they comply with the relevant environmental laws. The basic engineering design of landfills is to prevent the contamination of the ground and groundwater around the landfill. In essence, landfills for solid waste must be designed and constructed to safeguard the environment against contaminants that could be present in the waste stream.

SOLID WASTE DUMP
Solid waste dump site

Many of the concerns with landfills in the past were caused by poorly managed and improperly engineered dump sites. The disposal of waste in landfills has a lot of possible environmental consequences. The potential for groundwater and surface water pollution, the unchecked movement of landfill gas, and the generation of odour, noise, and visual nuisances are just a few of the long-term problems that may arise.

The dangers to human health resulting from the disposal of waste will be prevented, or at least reduced, to the greatest extent practicable, by proper landfill site design. It is important that the designer embrace practices, standards, and operational frameworks that are based on best practices currently in use and that take into account advancements in management practices and containment standards. The design approach should take into account the need to safeguard both human health and the environment.

Designing a landfill is a collaborative process that takes into account conceptual design ideas, results of environmental assessments and environmental monitoring, risk assessment, and findings from site investigations. Sustainable development is the main goal of waste management. Therefore, it is implied that landfill development and operation, which are inextricably intertwined, should take this strategy into account.

landfill construction
Construction of a landfill

In addition to providing additional precautions, the landfill siting plan limits the placement of landfills in environmentally sensitive locations while on-site environmental monitoring systems look for any indication of groundwater pollution and landfill gas. Additionally, a lot of modern landfills capture potentially dangerous landfill gas emissions and turn them into electricity.

The main goal of landfill site design is to offer efficient control measures to prevent or reduce, as much as possible, adverse effects on the environment, in particular the contamination of surface water, groundwater, soil, and air, as well as the resulting risks to human health resulting from the landfilling of waste.

The soil properties, geology, and hydrogeology of the site, as well as any potential environmental effects, all affect a landfill’s architectural idea. A site-specific design should be able to be created with the help of the studies for a landfill.

image 10
Figure 1: Cross-section of a typical modern sanitary landfill (Megooda et al., 2006)

The philosophy of landfill design has changed recently from the dry storage concept to the bioreactor approach. Leachate is recirculated in the bioreactor approach to increase the moisture content of the municipal solid waste and speed up biodegradation. This is a financially viable solution because it would be costly to dispose of collected leachate securely. By recirculating leachate, one can avoid the costly treatment cost of leachate.

In addition, waste degrades quickly as a result of the high moisture content brought on by leachate recirculation. Consequently, bioreactor landfills offer a significant decrease in post-closure management time and operation expense (Reddy and Bogner 2003).

A bioreactor landfill is described by SWANA (2001) as “any permitted landfill or landfill cell, subject to new source performance standards/emissions guidelines, where liquid or air, in addition to leachate and landfill gas condensate, is injected in a controlled manner into the waste mass to accelerate or enhance bio-stabilization of the waste.”

There are three different types of bioreactor landfills:

  • anaerobic,
  • aerobic, and
  • hybrid.

In anaerobic bioreactor landfills, anaerobic microorganisms (those that do not need oxygen for cellular respiration) speed up biodegradation. These bacteria turn organic wastes into organic acids, which are then converted into methane and carbon dioxide (Sharma and Reddy 2004). Aerobic microorganisms, which need oxygen for biological respiration and create carbon dioxide, are used in aerobic bioreactor landfills. Hybrid bioreactor landfills combine the aforementioned two methods.

Types of Landfills

The Environment Protection Agency (EPA) reports that landfills are controlled under RCRA Subtitle D (solid waste) and Subtitle C (hazardous waste), or by the Toxic Substances Control Act (TSCA).

States and local governments are responsible for the principal planning, regulating, and implementing bodies for the management of nonhazardous solid waste, such as domestic waste and nonhazardous industrial solid waste (Subtitle D);

Subtitle D landfills include the following:

Municipal Solid Waste Landfills (MSWLFs) – Specifically designed to receive household waste, as well as other types of nonhazardous wastes.

Bioreactor Landfills – A type of MSWLF that operates to rapidly transform and degrade organic waste.

Industrial Waste Landfill – Designed to collect commercial and institutional waste (i.e. industrial waste), which is often a significant portion of solid waste, even in small cities and suburbs.

Construction and Demolition (C&D) Debris Landfill – A type of industrial waste landfill designed exclusively for construction and demolition materials, which consists of the debris generated during the construction, renovation and demolition of buildings, roads and bridges. C&D materials often contain bulky, heavy materials, such as concrete, wood, metals, glass and salvaged building components.

Coal Combustion Residual (CCR) landfills – An industrial waste landfill used to manage and dispose of coal combustion residuals (CCRs or coal ash). EPA established requirements for the disposal of CCR in landfills and published them in the Federal Register April 17, 2015.

Subtitle C establishes a federal program to manage hazardous wastes from cradle to grave. The objective of the Subtitle C program is to ensure that hazardous waste is handled in a manner that protects human health and the environment. To this end, there are Subtitle C regulations for the generation, transportation and treatment, storage or disposal of hazardous wastes. Subtitle C landfills including the following:

Hazardous Waste Landfills – Facilities used specifically for the disposal of hazardous waste. These landfills are not used for the disposal of solid waste.

Polychlorinated Biphenyl (PCB) landfills – PCBs are regulated by the Toxic Substances Control Act. While many PCB decontamination processes do not require EPA approval, some do require approval.

landfill section
Typical section of a landfill

Design Considerations for Landfills

The designer should consider all environmental media that may be significantly impacted through the life of the landfill. The chosen design will have a major influence on the operation, restoration and aftercare of the facility. Aspects that must be considered in the design are briefly discussed below.

(1) Nature and quantities of waste
The waste types accepted at the landfill will dictate the control measures required. The requirements at a landfill accepting inert waste will be different to those at one accepting non-hazardous biodegradable waste which in turn will be different from a facility accepting hazardous waste.

(2) Water control
To reduce leachate generation, control measures may be required to minimise the quantity of precipitation, surface water and groundwater entering the landfilled waste. Contaminated water will need to be collected and treated prior to discharge.

(3) Protection of soil and water
A liner must be provided for the protection of soil, groundwater and surface water. The liner system may consist of a natural or artificially established mineral layer combined with a geosynthetic liner that must meet prescribed permeability and thickness requirements.

(4) Leachate management
An efficient leachate collection system may have to be provided to ensure that leachate accumulation at the base of the landfill is kept to a minimum. The leachate system may consist of a leachate collection layer with a pipe network to convey the leachate to a storage or treatment facility.

(5) Gas control
The accumulation and migration of landfill gas must be controlled. Landfill gas may need to be collected with subsequent treatment and utilisation, or disposal in a safe manner through flaring or venting.

(6) Environmental nuisances
Provisions should be incorporated in the design to minimise and control nuisances arising from the construction, operation, closure and aftercare phases of the landfill. Nuisances that may arise from landfilling include; noise, odours, dust, litter, birds, vermin and fires.

(7) Stability
Consideration must be given to the stability of the subgrade, the basal liner system, the waste mass and the capping system. The subgrade and the basal liner should be sufficiently stable to prevent excessive settlement or slippages. The hydraulic uplift pressure on the lining system due to groundwater must be considered. The method of waste emplacement should ensure stability of the waste mass against sliding and rotational failure. The capping system should be designed to ensure stability against sliding.

(8) Visual appearance and landscape
Consideration should be given to the visual appearance of the landform during operation and at termination of landfilling and its impact on the surrounding landforms.

(9) Operational and restoration requirements
The designer must consider the manner of site development and the necessary site infrastructural requirements during landfill operation and restoration. Landfill sites should be developed on a phased basis. Site infrastructure should include for the provision of; site accommodation, weighbridge, waste inspection area, wheelwash, site services and security fencing.

(10) Monitoring requirements
The designer should consider monitoring requirements at the design stage. These should be consistent with the requirements outlined in the Agency’s manual on ‘Landfill Monitoring’.

(11) Estimated cost of the facility
The designer should estimate the cost of the total project (construction, operation, closure and aftercare) from commencement to completion. This should include the costs of planning, site preparation and development works, operational works, restoration/capping works, landfill aftercare, and monitoring. Consideration should be given to the financing of the facility at the design stage in order to ensure that sufficient funds can be generated to fund ongoing and potential liabilities.

(12) Afteruse
The designer should consider the intended afteruse of the facility. It should be compatible with the material components and physical layout of the capping system, the surrounding landscape and current landuse zoning as specified in the relevant development plan.

(13) Construction
Environmental effects during construction must be considered. These may include noise from machinery, dust from soil excavation and soil placement, disturbance, traffic diversion, and avoidance of pollution by construction related activities.

(14) Risk Assessment
The design and engineering of a landfill should be supported by a comprehensive assessment of the risk of adverse environmental impacts or harm to human health resulting from the proposed development.

Conclusion

Modern landfills are well-engineered facilities which are situated, constructed, operated, and monitored in line with both federal and municipal laws. In the written word, there are three different kinds of landfills. Traditional dry landfills are the most popular choice. Dry landfills are being replaced by bioreactor landfills as a more environmentally friendly option. The newest entry on the list is sustainable landfills. Resources can be mined and refilled in sustainable landfills.

It is possible to think about landfills as a reliable and abundant source of materials and energy. This is widely recognized in the developing world, where waste pickers are frequently seen scouring the trash for useful stuff. Either landfilling is discouraged in underdeveloped nations or materials are recovered from landfills. In this framework, it is possible to see the idea of sustainable landfills as offering a universal remedy for waste disposal in both developed and developing countries.

Geotechnical Site Investigation

Any engineering or building structure must always require some site investigation. A complete analysis of the soil and groundwater conditions to a significant depth below the surface using boreholes and in-situ and laboratory tests on the materials encountered may be required, as well as a simple inspection of the surface soils with or without a few shallow trial pits.

The objectives of a geotechnical site investigation are to ascertain the conditions and properties of the soil, rock, and groundwater in the site, and to obtain extra pertinent information about the site.

ONGOING SITE INVESTIGATION
On-going site investigation

The type, size, and significance of the planned structure should be taken into account in the subsurface exploration program for a specific site. The number and depth of the necessary soil borings are determined by these criteria, which aid in the design of the site investigation program. Locating subterranean utilities should be part of the planning for a site investigation (i.e., phone, power, gas, etc.). As a result, many days before the planned site investigation, a local “call before you dig” service should be informed where available.

The significance and foundation configuration of the structure, the complexity of the soil conditions, and any knowledge that may be available regarding the behaviour of existing foundations on comparable soils all influence the scope of the investigation.

Categories of Site Investigation

Structures and earthworks are divided into three “geotechnical categories” according to Eurocode 7 (Geotechnical Design).

Geotechnical category 1 refers to light structures like single- or two-story buildings, low retaining walls, and buildings with column loads up to 250 kN or walls loaded to 100 kN/m. The qualitative investigations in this category can be restricted to verifying the design assumptions, at the latest, during the supervision of construction of the works, provided that the ground conditions and design requirements are known from prior experience and the ground is not significantly sloping. Visual inspection of the site, occasionally combined with inspection of small test pits, or sampling from auger borings are considered to be the main components of verification.

Conventional building types are included in category 2 structures on locations without anomalous dangers, unusually challenging ground conditions, or extraordinarily demanding loading conditions. This category includes conventional substructures such as retaining walls, bridge piers and abutments, excavations and excavation supports, rafts, piles, and shallow spread footings. Quantitative geotechnical data is needed, however standard testing techniques in the field and lab, as well as for analysis and design, are judged sufficient.

Structures in category 3 are those that are extremely massive, peculiar in nature, involve anomalous dangers, or have an unusually difficult ground or loading conditions. This category includes buildings that are located in seismically active regions.

The investigations essential for category 3 include any extra specialist research that may be required in addition to those thought to be sufficient for category 2. The processes and interpretations should be documented with references to the tests if specialized or unusual test procedures are necessary.

Site Investigations and Professional Practices

Buildings and engineering structures built upon deep excavations require extensive investigations. They offer vital information on the soil and groundwater conditions to contractors submitting bids for the work, in addition to information for foundation design. So, by collecting accurate and competitive bids based on an adequate understanding of the actual situation, money is saved.

If the cost of excavation work represents a sizeable portion of the overall project, a reputable contractor will not take a chance on it; instead, a comparably high sum will be added to the tender to account for the unforeseeable conditions. It follows the saying that “You pay for the borings whether you have them or not“.

An engineer doing a site investigation may hire local workers for hand auger boring or trial pit excavation, or they may hire a contractor for boring and soil samples. The boring contractor may send samples to his own lab or to a third-party testing facility if laboratory analysis is necessary.

EXCAVATION OF OPEN TRENCH
Excavation of trenches for site investigation

After that, the geotechnical engineer analyzes the soil mechanics for foundation design. Alternatively, the entire investigation could be handled by a specialized company with comprehensive capabilities for drilling, sampling, field and laboratory testing, and soil mechanics analysis.

If preferred, in-situ testing can take the place of laboratory testing. In any case, the engineer in charge of overseeing the field and laboratory work on a daily basis should keep the site investigation’s objective in mind and continuously evaluate the data in a manner similar to that used when writing the report.

The relevance of characteristics like weak soil layers, deep rock weathering, and sub-artesian water pressure can be explored in as much detail as may be necessary while the fieldwork is still on by avoiding the omission of important information in this way.

Whatever method the engineer chooses to conduct his site investigation, it is important that the people or groups doing the task are diligent and absolutely trustworthy. The engineer has a significant obligation to his employers to select a qualified organization and to satisfy himself through the field, lab, and office work inspections that the work has been completed accurately and thoroughly.

Information Required from a Site Investigation

For geotechnical categories 2 and 3 the following information should be obtained in the course of a site investigation for foundation engineering purposes;

  1. The general topography of the site, including surface configuration, adjacent property, the presence of watercourses, ponds, hedges, trees, rock outcrops, etc., and the accessibility for construction equipment and vehicles.
  2. The position of underground utilities such as sewers, water mains, cable television, and telephone lines.
  3. The overall geology of the region, with a focus on the primary geological formations that underlie the site and the potential for subsidence due to mining or other factors.
  4. The prior usage and history of the site, including any defects or failures of current or former buildings related to foundation issues and the potential for toxic waste contamination of the site.
  5. Any unique characteristics, such as the potential for earthquakes or environmental factors like flooding, seasonal swelling and shrinking, permafrost, or soil erosion.
  6. The accessibility and quality of locally produced building materials, including water for construction, building and road stone, and concrete aggregates.
  7. Information on normal spring and neap tide ranges, extreme high and low tidal ranges and river levels, seasonal river levels and discharges, tidal and river current velocities, wave action, and other hydrographic and meteorological data for marine or river structures.
  8. A thorough record of the soil and rock strata, groundwater conditions, and any deeper strata that may have an impact on the site’s circumstances in any way within the zones affected by foundation-bearing pressures and construction activities.
  9. The results of tests conducted in the field and laboratories on soil and rock samples relevant to the specific foundation design or construction issues.
  10. The results of chemical investigations performed on soil, fill materials, and groundwater to identify potential negative effects on foundation structures.
  11. The findings of chemical and bacterial investigations performed on contaminated soils, fill materials, and gas emissions to assess the potential health hazards.

Items (1) through (7) above can be obtained by a general site reconnaissance (the “walk-over” survey”) as well as through research of geological memoirs, maps, and other published records. Walking around the site area closely will often provide major clues about subsurface structures.

For instance, hidden swallow holes (sinkholes) in chalk or limestone formations are frequently revealed by sporadic depressions and noticeable irregularity in the ground surface; soil creep is shown by wrinkling of the surface on a hillside slope or leaning trees; abandoned mine workings are shown by old shafts or heaps of mineral waste; glacial deposits may be indicated by mounds or hummocks (drumlins) in a generally flat topography; and river or lake deposits by flat low-lying areas in valleys.

The existence of springs or wells and marshy terrain covered with reeds are surface indicators of groundwater (indicating the presence of a high water table with poor drainage and the possibility of peat). In the case of huge projects across wide areas, geological expertise should be sought.

Information on potential long-term changes in groundwater levels should be sought after. Pumping from deep mine shafts or ceasing groundwater extraction for industrial purposes can slowly raise groundwater levels over a large area.

Aerial photography is a useful tool for site inspections on large sites. Photographs may be taken from balloons, drones, or model aeroplanes. A great deal about a site’s topography and geomorphology can be learned through expert interpretation of aerial photos. A well-established science, geological mapping from aerial images is done by specialized companies.

Both outdated publications and old maps should be investigated because they may reveal the location’s prior uses and are especially helpful when looking into historically significant locations. Maps, memoirs, and images or photographs of a location in the past are frequently available at local libraries or museums. For information on buried services and coal mine workings in Britain, contact the Geological Survey and local authorities.

The list’s items (8), (9) (10), and (11) are obtained by boreholes or other subsurface research techniques, as well as through field and laboratory testing of soils or rocks. It is important to characterize soil types and consistency in accordance with accepted standards of practice. The British Standard Code of Practice Site Investigations, BS 5930, outlines the common descriptions and classifications of soils in Britain.

SOIL BORING
Soil boring

Subsurface Investigation Methods

Methods of determining the stratification and engineering characteristics of subsurface soils are as follows;

  • Trial pits
  • Hand auger borings
  • Mechanical auger borings
  • Light cable percussion borings
  • Rotary open-hole drilling
  • Wash borings
  • Wash probings
  • Dynamic cone penetration tests
  • Static cone penetration tests
  • Vane shear tests
  • Pressuremeter tests
  • Dilatometer tests
  • Plate bearing tests

Detailed descriptions of the above methods as used in British practice are given in BS 5930 Site Investigations. Brief comments on the applicability of these methods to different soil and site conditions are given in the sections below.

In most cases, geotechnical category 1 investigations use trial pits. For shallow foundations, they are helpful for assessing the quality of weathered rocks. The most reliable method for determining the stage of deposition and characteristics of filled ground is to use trial pits expanded to trenches.

trial pit
Trial pit

In soils that remain stable in an unlined hole, hand and mechanical auger borings are also suitable for category 1 examinations. Augering, when done properly, gives the least disturbance to the soil out of any other boring technique.

In British practice, light cable percussion borings are typically utilized. The straightforward and durable machinery is ideally suited to the vastly different soil types in Britain, including the extremely stiff or dense stone glacial soils and weathered boulders with a consistency similar to soil. For specialized testing, large-diameter undisturbed samples (up to 250 mm) can be recovered.

Typically, the United States, the Middle East, and countries in eastern Asia use rotary open-hole drilling. The rotary drills are typically skid- or tractor-mounted and can drill through rock as well as through dirt. Sample sizes are typically limited to 50 mm in diameter, and hole diameters are typically lower than percussion-drilled holes. Although drilling fluids such as bentonite slurry or water are employed, specialized foams have been created to aid in collecting nice, undisturbed samples.

Wash borings are holes with a small diameter (about 65 mm) that are bored using a water flush and chiselling. Sampling is done using 50 mm internal diameter open-drive tubes or 50–75 mm standard penetration test equipment.

Investigations into over-watered soils employ wash probings. They are used, for instance, in dredging investigations, to find rock heads or a strong layer overlain by loose or soft soils. They consist of a small-diameter pipe shot down. The soils cannot be positively identified, and sampling is frequently not feasible.

Soil Sampling

There are two main types of soil samples which can be recovered from boreholes or trial pits;

(a) Disturbed samples, as their name implies, are samples taken from the examples of the boring tools are auger parings, the contents of the split-spoon sampler in the standard penetration test, sludges from the shell or wash-water return, or hand samples dug from trial pits.

(b) Undisturbed samples, obtained by pushing or driving a thin-walled tube into the soil, represent as closely as is practicable the in-situ structure and water content of the soil.

RQD
Samples for rock quality designation (RQD)

It is important not to overdrive the sampler as this compresses the contents. It should be recognized that no sample taken by driving a tube into the soil can be truly undisturbed. Disturbance and the consequent changes in soil properties can be minimized by careful attention to maintaining a water balance in the borehole. That is, the head of water in the borehole must be maintained, while sampling, at a level corresponding to the piezometric pressure of the pore water in the soil at the level of sampling.

This may involve extending the borehole casing above ground level or using bentonite slurry instead of water to balance high piezometric pressures. The care in the sampling procedure and the elaborateness of the equipment depends on the class of work which is being undertaken, and the importance of accurate results on the design of the works.

Sinkhole Management in Construction Project Sites

The presence of well-developed solution channels, caves, springs, sinkholes, and extremely irregular weathered bedrock surfaces with cavities are the characteristics of a karstic topography. Sinkholes are “closed depressions” in the earth’s surface that are created when limestone and other rocks close to the surface dissolve and the material on top of them subside or collapses into subsurface solution caverns.

When infiltrating acidic water comes into prolonged contact with limestone, sinkholes often result. For obvious reasons, sinkholes are the major geologic hazard in karst terrain. They can cause structural damage and failure of buildings, cut off highways, drain ponds and lakes, and allow direct infiltration of contaminated groundwater.

solution caves

Sinkholes are the result of a slow, continuous process. However, the effects of the sinkhole near the surface can occur suddenly and catastrophically. There are two types of sinkholes:

i) Collapse Sinkhole, and
ii) Subsidence Sinkhole

Collapse Sinkholes
Collapse sinkholes develop when the limestone’s solution forms a vertical cavern or throat below the surface of the ground. Initially, the surface soil may be strong enough to span across the cavern. The bridging soils will eventually crumble as the cavern widens over time. Most people have heard about sinkholes in the media in terms of this.

Subsidence sinkholes
When the soil is relatively granular above the limestone formation, sinkholes might arise. In this situation, as the limestone erodes, the earth material on top fills the spaces left behind. This is known as ravelling, and if the earth keeps ravaging into the spaces in the limestone, the ground will eventually sink and produce a sinkhole.

The size of sinkholes can be influenced by the thickness of the overlying layer. The overlaying stratum can span a bigger cavern if it is sufficiently thick. For a collapse to occur, the cavern must enlarge. Additionally, sinkholes can grow to sizes large enough to engulf entire buildings.

sinkhole 2
Large collapse sinkhole

Site Investigation for Sinkholes

Rocks that are prone to solutioning activity are known to underlie around one-fourth of the earth’s crust (Syed, 2004). Numerous reports of issues with construction on karst sites have been made worldwide. Similarly, reports have been made on the sudden collapse of previously undetected solution cavities that have caused harm to existing structures and infrastructures.

It goes without saying that thorough subsurface investigations are required when significant structures are to be found in karstic regions. However, it is still true that it might be challenging to find possible sinkholes, caverns, or solutioning activity that have not yet affected the ground surface.

Because a borehole only examines a small region, the typical geotechnical study method of drilling holes may not be able to find them. Experience has shown that utilizing borings to locate sinkholes or cavities is only 10% to 20% accurate.

Studying the local geology and hydrogeology and mapping sinkholes that have already formed in the project area are the first steps in the investigation of sinkhole potential. Both surface geophysical approaches and boreholes will need to be used for the large-scale site investigation.

The typical approach should be to use borehole methods for detection and delineation and surface methods for early reconnaissance (anomaly detection). In other words, the zone in question should be drilled and sampled to give observations for the purpose of evaluation when aberrant responses are recorded during the surface surveys.

Various approaches to investigating Karstic features are:

• Aerial and Satellite Photography
• Backhoe Trenches
• Drilling boreholes
• Modern Geophysical Techniques

sinkhole problem
Sinkhole problem

Aerial and Satellite Photography

Identifying probable sinkholes can be done through aerial surveys. Large-scale zones of subsidence may be visible on old aerial photographs, which may aid in locating smaller, localized sinkholes. Although they can be located on the ground, an aerial survey is more practical.

Backhoe Trenches

Backhoes can easily and quickly explore a relatively large area. Trenches done this way can expose near-surface solution voids or sinkhole throats. However, they can not completely replace information obtained from a borehole.

Drilling Boreholes

Test borings are an important part of sinkhole or cavity investigations. The holes are drilled to the bedrock even if this requires drilling to much greater depths than would be necessary, otherwise. Standard Penetration Tests (SPTs) are usually conducted as the bore advances. This helps in knowing the strengths of various sediment layers. The data is used to draw subsurface cross-sections that help in inferring the presence or absence of sinkholes or cavity-associated features.

Geophysical Surface Surveys

A number of such techniques are currently in vogue that can locate cavities and sinkholes. The idea behind such techniques is to probe the subsurface without disturbing the ground surface. This is done by generating a wave, which when propagated through the soil, reveals anomalies. This can be investigated to find if the same is the presence of cavities or not.

Sinkhole Management in Construction – Case Study of Saudi Arabia

Syed (2004) published a case study about the Prince Abdullah Military City Project Site, which is in the oasis of Al-Hassa, in the Eastern Province of the Kingdom of Saudi Arabia, some 12 km west of the city of Hofuf. The report was published and presented during the fifth International Conference on Case Histories in Geotechnical Engineering, April 2004.

According to Syed (2004), there are several bedrock cavities at the project site. Numerous Reports, including those written in 1992 by a geotechnical consultant for a small area of the subject Site and those completed by the US Army Corps of Engineers in 1978, have documented these conclusions for the subject site.

The following terms best describe the karstic features seen at the project site: collapse sinkholes, subsidence sinkholes, dropouts, and bedrock cavities. In order to find and map the bedrock cavities at the site, a rigorous soil investigation program was launched. In-situ testing and sampling of overburden soil by Standard Penetration Tests (SPTs) and coring in the bedrock strata were all parts of the detailed geotechnical investigation program that was carried out at the site for the purpose.

Additionally, Core Recovery and Rock Quality Designation (RQD) measurements were made. Testing typical samples of subsurface materials in the lab was also included. To further confirm the competency of the stratum below, a thorough “Cavity Search Probing” was also carried out under the footprints of each building. With the aid of a wagon drill and “Pneumatic Driving,” a rock probe was driven into the bedrock while being cleaned out of the hole as it went.

The “Karstic Terrain” at the project site was mapped with the aid of these investigation and probing. The information was also helpful in developing plans for the project’s remediation and later construction. Given the karstic issues at the project site, a semi-rigid raft foundation was chosen. This foundation system became the best option for the situation because it is known for being quite robust to bridge over the underlying cavities.

Sinkhole Investigation at the Site

The Project Site spans an area of around 6 square kilometers and is divided into 8 zones. An initial preliminary reconnaissance investigation was conducted at the site, followed by a detailed investigation for each of the 8 Zones. There were already two investigations conducted at the project site prior to the preliminary reconnaissance investigation (conducted in March 1998), each conducted by a different agency at a different time. The range of work for the various investigations carried out at the project site can be found in Syed (2004).

All of these tests at the project site proved that the limestone that makes up the bedrock is primarily light brown, fine-grained, strongly to moderately worn, and jointed. The underlying rock strata’s Total Core Recovery (TCR), which generally exceeds 50%, was determined to be between 27% and 100%. However, the Rock Quality Designation (RQD), which ranged from 0% to 100%, was typically lower than 20%.

The presence of several small cavities inside the underlying limestone bedrock is one of this limestone’s common characteristics that has been mentioned in all the studies. The findings advised performing a thorough Cavity Probing after excavation during the building stage at the locations of the foundations (under the footprints of the facilities). Drilling probing holes at least 1.5 times the width of the foundations deep is the best way to accomplish this.

The following methods were applied at the project site to investigate the local geology and hydrogeology as well as to find and map probable sinkholes and cavities.

• Trenches and Test Pits
• Drilling Boreholes, and
• Cavity Probing using pneumatic driving of Probe.

Trenches, test pits, and boreholes were used to collect a lot of information that indicated the presence of bedrock cavities beneath the surface. However, before starting to pour the foundations, a thorough cavityprobing was done beneath the footprints of each facility to determine whether or not there were any cavities. 5,610 probe holes total (in 166 facilities in 6 zones) were made. This comprised an extra 219 numbers in 18 numbers of facilities located surrounding the problematic locations.

The Cavity search program was quite thorough and detailed. The probing was done by using a wagon drill to pneumatically drive a rock probe into the rock. As time went by, compressed air was utilized to clear the hole. The probe’s time through each successive 20 cm depth was meticulously measured to determine the rock’s resistance to penetration.

image 8
Cavity Probing in progress at the Site (Syed, 2004)

The time for penetration records will be quite short if there are cavities or loose zones in the underlying strata. For a 20 cm penetration, a time of less than 10 seconds is regarded as the presence of loose zones, however in the case of a cavity, there will be no resistance to rock penetration and it happens suddenly.

In the course of this operation, the following details were meticulously recorded:

• Time taken for 20 cm penetration
• Air escape in the finished boreholes
• Sudden fall of drill rod
• Boreholes, where time was 10 seconds or less

Any bore where the previous three observations were made were probed repeatedly until a cavity was identified or some logical explanations for the observations were found. If there were any cavities, grout was used to fill them in. To reveal the sinkholes or cavities, excavations were also carried out in the vicinity of these problem probing holes.

Treatment and Remedial Measures

The methods used to remedy Karst-related features may be as diverse as the cavities and sinkholes themselves. In order to expose the opening in the rock surface, the overburden soils must be removed if the “throat” of a sinkhole can be found.

The excavation is then backfilled once the throat has been sealed off or covered with an inverted filter. A less effective treatment is typically employed if the throat cannot be found or if the depth of the rock makes exposing the rock surface impractical.

image 9
Typical sinkhole at the site (Syed, 2004)

The excavation base could subsequently be sealed with concrete and/or the entire region could be covered with a geotextile if a throat is discernible but the depth to rock is excessive. If there is no neck visible and the rock is too deep, the only practicable remedy is to remove soft or organic materials, cover the region with a geotextile, and then re-fill the hole with clayey soils.

The project site’s sinkhole and cavity issues, however, required a variety of treatments and corrective actions due to the unique site characteristics.

This can be grouped as:

i) during leveling and grading
ii) during foundation construction, and
iii) during external works construction, like roads, water supply, sanitary sewer, storm water drainage, and other items related to landscaping and irrigation works.

With the exception of a few, the majority of the sinkholes and cavities discovered during foundation construction were outside the footprints of the facilities. Large open-mouthed sinkholes of about 5m × 4m x 3m deep were discovered inside the footprint of some foundation areas. These were overexcavated to reveal the size of the cavities, and then compaction grouting was used to cure them. This innovative method, which is known as “the Grout injected with a slump less than 25 mm,” was developed in the USA.

Typically, a soil-cement mixture with large amount of silt particles to give plasticity and create internal friction is utilized in this. In most cases, grout does not penetrate soil pores; instead, it stays in a homogenous mass that allows for controlled displacement. Additionally, regular grout was employed to address the issue in this facility. Some of them were also backfilled with rock using a bridging technique.

Due to the short duration observed for a 20 cm probe penetration during a cavity search probe, there was suspicion of the presence of sinkholes or cavities in some other facilities. Therefore, more investigations were conducted to confirm the suspicions. Except for two sites, the time recorded for the additional probes in most cases did not record fewer than 10 seconds.

Deep excavations have to be done for the service pits and hydraulic lift pits at some of the facilities. These extra probing were nearby and had time records of less than 10 seconds, which suggested the presence of cavities. Therefore, it was determined to over-excavate the limited area in order to explore the underground caverns.

The cavities were exposed during excavation, along with several lateral solution channels. These were completely cleaned before being treated using two different techniques:

(i) Compaction grouting of the vertical cavities, and
(ii) Bridging utilizing boulders and geotextiles to seal the mouth of the lateral solution channels.

Foundation Design Considerations

Even with the best techniques and designs, construction in karst terrain is unquestionably not risk free, according to experts on the issue. However, the risk they pose justifies the search for a remedy that lowers the risks at a reasonable cost.

The solution in vogue are outlined as follows;

• Optimize the location on the site
• Treat defects that are present
• Use modified shallow foundations
• Use deep foundations
• Minimize future activation

For medium-sized buildings, the use of modified shallow foundations is a practical solution to the problems of Karstic hazards.

Utilizing such foundations involves:

(i) constructing a footing that spans or bridges over the cavity; and 
(ii) constructing a mat foundation that is rigid enough to minimize deflections that may occur due to Sinkhole formation beneath it.

The proposal accepted for the project site is a semi-rigid raft foundation. It has thicker borders or raft bands and resembles a mat. This was implemented in the project to get around the site’s sinkhole and cavern issues. This can span the cavities because it is sufficiently stiff. The same was analyzed and designed using PCA-MATS. A bearing capacity of 150 kN/m2 was used to design the foundations in accordance with the Geotechnical Report. Given the rock’s worn state and the karstic terrain, a relatively low figure of bearing capacity was advised.

Conclusion

Sinkholes are geologic hazards that can pose enormous risk to the safety of infrastructures. Extensive geotechnical site investigation is required when there is a possibility of the existence of sinkholes and cavities in an area. In a report presented by Syed (2004), a Karstic terrain was encountered in a project site in Al Hofuf, Saudi Arabia.

The detection, delineation, and mapping of bedrock cavities, sinkholes, dropouts, and solution channels were the subject of a thorough investigative program. Under the footprints of 166 facilities in 6 zones, 316 boreholes, 10 test pits, and 5610 cavity probe holes were drilled. By employing a Wagon Drill to pneumatically drive a rock probe, the cavity was probed. As time went by, compressed air was utilized to clear the hole.

The site’s karstic features were mapped, and remediation efforts were made. The majority of the sinkholes and voids were filled in using grout, with the use of geotextiles in other locations. A novel method of compaction grouting was used to fix some of them. Some of them were also “plugged” with rock utilizing the bridging method. By creating masonry and/or concrete structures, the voids and lateral solution channels in utility trenches were sealed. The roadways’ ditches were paved.

The Site’s treated karstic features are functioning effectively, and for the past four years or more, no problems have been reported. A hard and shallow style of foundation called a semi-rigid raft foundation was employed. Such footings are suggested by experts because of their capacity to span the cavities.

Reference Article:
Syed, Ahmad Faiz, “Managing Sinkholes at Project Site, A Saudi Arabian Case History” (2004). International Conference on Case Histories in Geotechnical Engineering. 20. https://scholarsmine.mst.edu/icchge/5icchge/session06/20

Groundwater Control: Exclusion Techniques

Many civil engineering projects usually involve excavating into water-bearing soil formations. Before beginning such excavations, the proper system(s) for managing and controlling groundwater and surface water run-off should be planned. This can only be done in practice if you are aware of the ground and groundwater conditions you are likely to face through site investigation data.

It is necessary to take precautions to manage groundwater flows and pore water pressures in water-bearing soils in order to prevent problematic circumstances during excavation and construction. Effective management of surface water runoff is also necessary. Understanding the potential effects of an excavation can help determine which groundwater control measures are required to assure stability.

There are three groups of methods available for temporary works control of groundwater:

(a) Lowering groundwater levels in construction by means of water abstraction, in other words – groundwater lowering or dewatering.
(b) Exclusion of groundwater inflow to the area of construction by some form of very low permeability cut-off wall or barrier (e.g. sheet-piling, diaphragm walls, artificial ground freezing).
(c) Application of fluid pressure in confined chambers such as tunnels, shafts and caissons to counterbalance groundwater pressures (e.g. compressed air, earth pressure balance tunnel boring machines).

Groundwater Control Techniques

Techniques for the control of groundwater can be divided into two principal types:

(a) Those that exclude water from the excavation (known as exclusion techniques)
(b) Those that deal with groundwater by pumping (known as dewatering techniques)

Exclusion Methods of groundwater control

The aim of groundwater control by exclusion is to prevent groundwater from entering the working area. Creating a physically impermeable cut-off wall or barrier around the perimeter of the excavation to keep groundwater out is one of the most widely used exclusion methods. The cut-off often produces a basal seal for the excavation by penetrating vertically down to a very low permeability stratum.

image 7
Exclusion method of groundwater control

The depth and make-up of any underlying permeable stratum have a significant impact on the costs and viability of constructing a physical cut-off wall. Base instability is possible when upward seepage occurs beneath the base of the cut-off wall due to the lack of or presence of a sufficient very low permeability layer. In these situations, dewatering techniques and exclusion techniques may be combined.

As an alternative, a horizontal barrier or “floor” might be formed adjacent to the cut-off structure to stop vertical seepage. Although it is not common, horizontal barriers have been built utilizing techniques such jet grouting, mix-in-place grouting, and artificial ground freezing.

A portion of the groundwater will become trapped inside the working area if a complete physical cut-off is established. In order to move further with the project, this must be removed, either by sump pumping during excavation or by pumping from wells or wellpoints beforehand.

The capacity of the exclusion technique to enable work to be carried out below the groundwater level with little impact on groundwater levels outside the site is one of its attractive features. This ensures that any groundwater-lowering problems are prevented. Exclusion techniques are frequently employed instead of dewatering techniques, especially in metropolitan areas, to reduce the danger of settlement damage brought on by reducing groundwater levels.

However, it is crucial to remember that practically all walls will leak to some degree when considering the use of the exclusion technique to prevent groundwater level from dropping in areas beyond the site. Particularly vulnerable to leakage are any joints (between columns, panels, piles, etc.) left behind from the installation process.

Several issues can arise if groundwater leaks through cut-offs into the excavation or work area:

(i) The seepages may hinder site operations during construction, necessitating the deployment of sump pumps or surface water management techniques to keep the working area dry.
(ii) The risk of settlement or other negative impacts may result from the leaking into the excavation being severe enough to locally lower groundwater levels outside the site.
(iii) If the cut-offs are a permanent construction, like the walls of a deep basement, even tiny seepages over time will be ugly and could interfere with any architectural finishes that have been applied to the walls.

Grouting or other treatments can frequently be used to address the major seepages that cause issues (i) and (ii). On the other hand, it might be quite challenging to stop or stop the little seepages of (iii). Costs significantly increase if a completely dry or leak-proof construction is required. If cut-off walls are going to be included in the permanent works, this is something that needs to be taken into account.

The techniques used in groundwater exclusion are listed below;

Steel Sheet Piling

In this groundwater control technique, steel sheet piles are driven into the soil to form a barrier against the intrusion of groundwater into the construction area. This is one of the most prominent techniques used in the construction of cofferdams. This technique is applicable to most open soils, however, it can be challenging when obstructions such as rock boulders are encountered.

cofferdam in the world

Steel sheet pile walls may be installed to form permanent cut-off, or used as temporary cut-off with piles removed at the end of construction. They offer the advantage of rapid installation and are relatively cheap. Additionally, they can support the sides of the excavation with suitable propping.

The disadvantages are that the seal may not be perfect, especially if obstructions are encountered. Vibration and noise of driving may be unacceptable on some sites, but ‘silent’ methods are available.

Vibrated beam wall

In this method, a grout injection nozzle is driven into a specially made wide flange beam section using a vibratory driver-extractor that is attached to the beam’s base. A self-hardening slurry is injected into the ground with the designed beam while it is vibrated into the ground to act as a lubricant.

When the beam is withdrawn, the extracted beam element leaves a minimum of a 4 to 6-inch panel that is filled with the self-hardening slurry. A continuous cutoff wall is created by the consecutive beam element insertions and the overlapping of the earlier beam insertions.

vibrated beam cutoff wall
Vibrated beam cut-off wall

The Vibrating Beam construction technology enables operations in small spaces with little room for above-ground mixing or staging. This method also decreases soil disposal costs, which can be expensive when dealing with hazardous locations, as less excavation is needed. Permeabilities in the range of 10-8 cm/sec are capable of allowing for depths greater than 50 ft.

This method is applicable in open excavations in silts and sand but will not support the soil.

Slurry trench cut-off wall

A slurry cut-off wall or slurry trench wall is an excavation made deep into the ground while simultaneously pumping an engineered slurry mix into the trench. A permanent low permeability barrier to groundwater and leachates is created by the slurry cut-off wall after it is completed. Slurry trench cut-off walls also have the capacity to prevent the transportation/movement of a range of heavy metals and organic contaminants including volatile organic compounds (VOCs), hydrocarbons, diesel, solvents and tar.

slurry wall
Slurry trench cut-off wall

The slurry trench walls create a low permeability curtain wall surrounding the excavation and it is a permanent water exclusion system. Slurry trench walls are can be quickly installed and relatively cheap, but the cost increases rapidly with depth. It is very applicable in silts, sands, and gravel up to a permeability of about 5 x 10-3 m/sec.

Structural Concrete Diaphragm wall

Diaphragm walls can be used as cut-off walls for dams or excavation pits, as foundations, or as enclosures for structures. A diaphragm wall is a structural concrete wall constructed panel by panel in a deep trench excavation using either precast or in-situ concrete pours.

Diaphragm walls can serve as retaining walls, water-cut-off structures, deep foundations, basement walls, or as separating structures for underground facilities. They are constructed as ground-level concrete or concrete reinforced with steel walls. They are thought to be almost water-impermeable and deformation-resistant.

image 6
Diaphragm wall in shaft construction

Diaphragm walls are permanent structures that support the sides of the excavation and often form the sidewalls of the finished construction. They can be keyed into rock and have the advantage of minimum noise and vibration. However, high cost may make the method uneconomical unless walls can be incorporated into a permanent structure.

It is applicable in side walls and shafts in most soils and weak rocks but the presence of boulders may cause problems.

Secant (interlocking)and contiguous bored piles

A retaining wall made of closely spaced bored piles can be used to construct a deep basement or a cut-and-cover tunnel. The piles could be constructed so that they almost touch each other (contiguous). A watertight retaining wall can be created by grouting the spaces between the piles.

secant piles
Secant piles

This method of construction results in a more effective form of structure when the piles interlock. The piles will typically need propping during soil excavation before the permanent floor and/or roof structure is finished.

Jet Grouting

In jet grouting, a stabilizing fluid is injected into the subsoil (or the soil being treated) under high pressure and high velocity. The high-velocity fluid jets are used to create different geometries of cemented soil in the ground and typically form a series of overlapping columns of soil/grout mixture that can prevent the movement of groundwater.

Jet grouting can be messy and create large volumes of slurry. There is a risk of ground heave if not carried out with care. Jet grouting is Relatively expensive and applicable for open excavations in most soils and very weak rocks.

Deep soil mixing columns

In this permanent water exclusion solution, overlapping columns are formed by in-situ mixing of soil and injected grout/cement using auger-based equipment. Soil mixing can be categorised into deep soil mixing and shallow soil missing. The mixture of binding agent and natural soil produces columns with very low permeability. This approach produces little spoil, and it is less flexible than jet grouting. It is relatively expensive.

deep soil mixing equipment

Injection grouting using cementitious grout

Injection grouting is a process by which grouts are injected under pressure into open fissures, voids, cracks, and pores in a soil/rock mass. The grout fills the pore spaces, preventing the flow of water through the soil.

injection grouting in a tunnel
Injection grouting in a tunnel

Equipment is simple and can be used in confined spaces. A comparatively thick zone needs to be treated to ensure a continuous barrier is formed. Multiple stages of treatment may be needed. The procedure is applicable to tunnels and shafts in gravels and coarse sands, and fissured rocks.

Artificial ground freezing

Ground freezing is a temporary water exclusion solution during construction. In this solution, a ‘wall’ of frozen ground (a freeze wall) is formed using brine or liquid nitrogen, which can support the side of the excavation as well as exclude groundwater.

ground freezing in construction
Ground freezing in construction

This approach may not work if groundwater flow velocities are excessive (> 2 m/day for brine or 20 m/day for liquid nitrogen). Liquid nitrogen is expensive but quick; brine is cheaper but slower. Liquid nitrogen is to be preferred if groundwater velocities are relatively high. Plant costs are relatively high.

Compressed air

In this temporary groundwater control solution, increased air pressure (up to 3.5 bar) raises pore water pressure in the soil around the chamber, reducing the hydraulic gradient and limiting groundwater inflow. Potential health hazards to workers. Air losses may be significant in high-permeability soils. High running and set-up costs.

It is suitable for confined chambers such as tunnels, sealed shafts and caissons.

Design of Light Gauge Steel Columns | Cold-formed Steel Columns

Axial compression loads from light gauge framed buildings must frequently be carried by light gauge steel members (cold-formed sections), such as the studs in a load-bearing wall. Similar to their hot-rolled counterparts, light gauge steel compression members’ failure is likely to be caused by buckling rather than cross-sectional yielding, yielding a member resistance that is much lower than the squash load of the section.

Since its buckling resistance must be calculated, the design process for such a member is in many ways comparable to that of hot-rolled steel columns. However, there are a number of ways in which the behaviour of light steel wall studs differs from that of hot-rolled columns, and these variations must be taken into account during the design process.

Contrary to columns, which function as separate parts inside a structural frame, load-bearing panels are created using light steel wall studs, plasterboard, and often some type of sheathing board. A certain amount of lateral constraint in the minor axis of the studs will be provided by the presence of the boards, which can be used to determine the buckling resistance. Any constraint must, however, be tested using studs of a representative slenderness range and a build-up of boards that is comparable to that used in actual practice.

light gauge framed building
Light gauge framed building

While hot-rolled steel columns typically behave according to flexural buckling, many light steel sections can also buckle in a torsional-flexural manner. This form of failure will naturally control the member’s resistance if torsional-flexural buckling occurs at a lower magnitude of load than flexural buckling.

The elastic critical buckling load utilized for design is assumed to be the least significant of the elastic critical buckling loads for flexural buckling, torsional buckling, and torsional-flexural buckling. This is reflected in the Eurocode design guidelines.

Last but not least, light steel sections are prone to local and distortional buckling, both of which can negatively affect a member’s ability to withstand compression. When estimating the compression resistance, this should be taken into account by utilizing the effective cross-sectional area rather than the area of the gross cross-section.

Design Procedure According to the Eurocodes

Clause 6.2 of BS EN 1993-1-3 outlines the design processes for compression members made of light gauge steel. But because the design of hot-rolled columns is comparable, designers are directed to Clauses 6.3 of BS EN 1993-1-1 for the majority of the details, including the buckling curves.

When a light gauge steel member is subjected to axial compression, the design buckling resistance is given by:

Nb,Rd = χAefffyM1

Where;
χ is the reduction factor for flexural buckling
Aeff is the area of the effective cross-section
fy is the yield strength of the steel
γM1 is the partial factor of safety for buckling

The reduction factor χ is used to quantify the reduction in resistance below the squash load of the section due to buckling. It may be obtained from BS EN 1993-1-1 using the appropriate buckling curve and the value of slenderness λ corresponding to the critical mode of failure.

BS EN 1993-1-1 offers a choice of 5 buckling curves, but this is restricted to 3 curves for light gauge steel according to Clause 6.2.2 of BS EN 1993-1-3. The appropriate choice of curve for various types of cross-section is given in Table 6.3 of BS EN 1993-1-3.

The slenderness λ is given by:

λ = √(Aefffy/Ncr)

Ncr is the elastic critical buckling load, which for flexural buckling is equal to the Euler load and is given by:

Ncr = π2EI/Lcr2

where:
E is Young’s modulus for the material.
I is the appropriate second moment of area (for the gross cross-section).
Lcr is the effective length between points of restraint.

The reduction factor χ may be obtained directly from the buckling curves printed in BS EN 1993-1-1 or from the following equations:

Φ = 0.5 [1 + α(λ – 0.2) + λ2]
χ = 1/[Φ + √(Φ2 – λ2)]

α is the imperfection factor corresponding to the chosen buckling curve. Values of α are given in Table 6.1 of BS EN 1993-1-1.

LIGHT GAUGE FRAMING

Design Example of Light Gauge Steel Columns

A light gauge steel building is to be constructed using cold-formed sections. Determine the buckling strength of a steel column in the wall stud constructed using a lipped channel section (200 x 65 x 2) which is restrained at the top at 3.5 m in the y-direction, and at 1.75m in the z-direction.

Length of member between restraints:
Ly = 3.50 m
Lz = 1.75 m

Effective lengths (assuming that the member is pin-ended):
Ly,cr = 3.00m
Lz,cr = 1.50m

Section depth h = 200 mm
Flange width b = 65 mm
Stiffener depth c =25 mm
Corner radius r =3 mm
Nominal thickness tnom = 2 mm
Core thickness t = 1.96 mm
Design strength fy = 350 N/mm2
Young’s modulus E = 210000 N/mm2
Partial safety factor γM1 = 1.00

Gross section properties
Area of gross cross-section A = 729 mm2
Second moment of area (major axis) Iy = 440.5 cm4
Second moment of area (minor axis) Iz = 44.26 cm4

Effective section properties
The effective area of the cross-section in compression: Aeff = 459.1 mm2 (This effective area has already been calculated by Heywood and Way, 2012).

Along the major axis;
Lcr,y = 3.5 m
Ncr = π2EI/Lcr2
Ncr = (π2 × 210000 × 4405000)/35002 = 745296.1266 N
λ = √(Aefffy/Ncr) = √(459.1 × 350)/745296.1266 = 0.464

Along the minor axis;
Lcr,z = 1.75 m
Ncr = π2EI/Lcr2
Ncr = (π2 × 210000 × 442600)/17502 = 299539.6737 N
λ = √(Aefffy/Ncr) = √(459.1 × 350)/299539.6737 = 0.732

The buckling curve b is appropriate for the y-y and z-z axis. The imperfection factor for buckling curve b, α = 0.34

Φ = 0.5 [1 + α(λ – 0.2) + λ2]

Φy = 0.5 [1 + 0.34 (0.464 – 0.2) + 0.4642] = 0.652
Φz = 0.5 [1 + 0.34 (0.732 – 0.2) + 0.7322] = 0.858

X = 1/(Φ + √(Φ2 – λ2))
Xy = 1/[0.652 + √(0.6522 – 0.4642)] = 0.900
Xz = 1/[0.858 + √(0.8582 – 0.7322)] = 0.765

Therefore;
Nb,Rd,y = (Xy Aeff.fy)/γm1 = (0.9 × 459.1 × 350) / (1.0) = 144616.5 N = 144.616 kN
Nb,Rd,= (Xz Aeff.fy)/γm1 = (0.765 × 459.1 × 350) / (1.0) = 122924 N = 122.92 kN

In this case, the lesser holds for the flexural buckling resistance.

Hence Nb,Rd = 122.92 kN

Article reference:
Heywood M. and Way A. (2012): Design of light gauge steel elements. In Steel Designer’s Manual (Davison B. and Owens G. W. eds). Wiley-Blackwell,UK

Plate Girders

Plate girders are built-up beam sections manufactured to support massive vertical loads over long spans with consequent bending moments that are greater than the moment resistance of readily accessible rolled sections. The plate girder is a built-up beam that is made up of two flange plates that are fillet welded to a web plate to create an I-section (see Figure 1).

The primary function of the top and bottom flange of plate girders is to withstand the axial compressive and tensile forces induced by the applied bending moments, while the primary function of the web is to withstand shear. In fact, some codes of practice employ this division of structural action as the basis for design.

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Figure 1: Elevation and cross-section of a typical plate girder

By extending the distance between them, the required flange areas for a given bending moment can be decreased in a plate girder. Therefore, it is advantageous to increase the distance between flanges for an economical design. The web thickness should be decreased as the depth increases to minimize the girder’s self-weight, however, this makes plate girders more susceptible to web buckling issues than rolled beam sections.

Plate girders are frequently employed in small- to medium-span bridges and occasionally utilized in buildings as transfer beams. Design guidelines are in BS EN 1993-1-5 (2006). This article describes current plate girder design procedures and includes references to the pertinent code provisions.

plate girder bridge
Figure 2: Old plate girder bridge

Advantages and Disadvantages of Plate Girders

Plate girder fabrication prices have decreased significantly as a result of the advent of highly automated workshops, but box girder and truss fabrication costs are still high because these structures must still be made largely by hand.

When compared to rolled sections, fabricated plate girders can make better use of the material since the designer has more ability to alter the section to account for variations in the applied forces. As a result, plate girders with variable depth have become more prevalent in recent years.

Compared to trusses, plate girders are more aesthetically beautiful and easier to transport and build than box girders.

The use of plate girders has some limitations. They are heavier, more difficult to transport and have higher wind resistance than trusses. It is also more challenging to provide openings for building services. Because compression flange stability is an issue, plate girder erection can occasionally present challenges.

Selection of cross-section for plate girders

During the preliminary design of plate girders, some factors can be considered for the initial selection of the dimensions. They are described as follows;

Span-to-depth ratios

Modern fabrication techniques enable the cost-effective production of plate girders with constant or variable depth. Constant-depth girders have historically been more prevalent in buildings, but as designers show a greater willingness to alter the steel structure to accommodate services, this could change. Table 1 lists suggested ratios of span to depth for plate girders in buildings.

ApplicationsSpan-to-depth ratio
(1) Constant-depth beams used in simply-supported composite girders, 12 to 20 and for simply-supported non-composite girders with concrete decking12 – 20
(2) Constant-depth beams used in continuous non-composite girders using concrete decking (N.B. continuous composite girders are uncommon in buildings)15 – 20
(3) Simply-supported crane girders (non-composite construction is usual)10 – 15
Table 18.1: Recommended span-to-depth ratios for plate girders used in buildings

Recommended plate thickness and proportions

Although more slender cross-sections are permissible, in general, the slenderness of plate girders used in buildings should not exceed the restrictions established for Class 3 cross-sections (given in Clause 5.5 of BS EN 1993-1-1). The choice of plate thickness and cross-sectional buckling are connected. In order to restore proper stiffness and strength, if the plates are too thin, stiffening may be necessary; however, this additional labour is expensive.

In view of the above, the maximum depth-to-thickness ratio of the webs (cw/tw) of plate girders in buildings is usually limited to:

cw/tw < 124ε = 124(235/fyw)0.5

where fyw is the yield strength of the web plate.

The outstand width-to-thickness ratio of the compression flange (cf/tf) is typically limited to;

cf/tf < 14ε = 14(235/fyf)0.5

where fyf is the yield strength of the compression flange. Note that cw and cf are the flat element widths, measured from the edges of the fillet welds (or root radii for rolled sections).

For initial design purposes, when the weld size may be unknown, it is conservative to ignore the weld and take cw = hw (the distance between the flanges) for webs and cf = b/2 – tw/2 for outstand flanges. Changes in flange size along the girder are not usually worthwhile in buildings.

For non-composite girders the flange width is usually within the range 0.3 – 0.5 times the depth of the section (0.4 is most common). For simply-supported composite girders these guidelines can still be employed for preliminary sizing of compression flanges. The width of tension flanges can be increased by 30%.

Stiffeners

Plate girders used in structures don’t typically require longitudinal web stiffeners. When the resistance of the unstiffened web would otherwise be surpassed, transverse (vertical) web stiffeners may be used to increase the resistance to shear close to the supports or to carry highly concentrated transverse forces acting on the flanges. Because there is less shear in these areas, the necessity for intermediate stiffening reduces.

stiffener in plate girder
Figure 3: Stiffeners in plate girders

The use of intermediate transverse web stiffeners improves the web panels’ ultimate shear buckling strength τu (including post-buckling or post-critical), as well as elastic shear buckling strength τcr. A decrease in the web panel aspect ratio a/hw (width/depth) greatly enhances the elastic shear buckling strength.

Enhancing tension field action causes the boundary members to resist diagonal tensile membrane stresses that develop during the post-buckling phase, increasing the ultimate shear buckling strength (transverse stiffeners and flanges).

Since there is little gain in strength for smaller panel aspect ratios, intermediate transverse stiffeners are often positioned such that the web panel aspect ratio is between 1.0 and 2.0. In order to create what is known as a rigid end post, pairs of stiffeners are occasionally used at the end supports. Generally speaking, the girder’s overhang beyond the support is limited to no more than one-eighth of its depth.

Effects of Soil Structure, Water Content, and Density on the Expansion Potential of Soils

The level of risk associated with foundation or slab movement at a site is determined by the expansion potential of soil or sedimentary bedrock formation. Any soil or rock element that has the ability to expand in volume with increasing water content is often referred to as expansive soil.

The volume of soil decreases as a result of consolidation, which is a process that forces water out of the pores and fills them with soil particles. However, during swelling, water is absorbed into the soil, which is the reverse of consolidation, in that it forces the soil’s particles apart and causes an increase in volume. This swelling may result in issues very similar to differential settlement by lowering the soil’s ultimate bearing capacity and shear strength.

IMG 2094
Figure 1: Expansive soils can cause problems to structures and foundations

Swell and swelling pressure can result from elastic restitution after a load is removed, from water adhering to the surface of soil particles, or from the particles expanding as a result of the adsorption of water into the soil particles. Of these, the adsorption of water is the most common. To comprehend the phenomenon of adsorption, a basic understanding of clay mineralogy is required.

The soil structure, dry density of the soil, the initial moisture content, and the availability and characteristics of water are all factors that affect swelling properties in addition to soil composition.

Effect of Soil Structure on Expansion Potential of Soil

The way soil particles interact will depend on their orientation within the soil mass and their distance from one another. The particles may obtain different degrees of orientation depending on the circumstances present during the deposition.

The particle orientation for flocculated and dispersed soil structures is shown in Figure 2. The figure exemplifies the extremes of totally flocculated and fully dispersed soil structures. Particle orientation in most soils would fall between these two extremes. It is convenient to think about the structures that are exhibited while considering the effects of density and soil structure on expansion potential.

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Figure 2: Sediment structures: (a) flocculated orientation; (b) dispersed orientation

It is obvious that the interactions between micelles in a flocculated structure, such as the one in Figure 2a, are principally driven by the contacts between the ends of the particles and the faces of adjacent ones. In contrast to the dispersed structure, the flocculated structure has a greater gap between particles.

It follows that in contrast to the dispersed structure shown in Figure 2b, crystalline and osmotic swelling would be less effective in the flocculated structure. The soil structure would lean more toward the dispersed structure for highly overconsolidated clays that have been subjected to substantial overburden stresses.

Effects of Water Content on Expansion Potential

Swelling won’t occur until there is free water available. Studies have revealed that depending on the type of ions in the water and whether or not there are ions already adsorbed on the soil particles, the electrolyte concentration of the water may change the swelling characteristics. The soil’s expansive properties could be significantly changed if the ions in the water have the capacity to replace the ions that have been adsorbed there.

The expansion potential will also be influenced by the distance between particles, the hydration states of the cations, and particle orientation. As a result, it would be expected that soil with a high dry density and low initial water content would have a higher expansion potential than soil with a lower dry density and a higher initial water content.

Chen (1973 and 1988) conducted oedometer experiments on samples that had been compacted to the same initial density but with different initial water contents. The results of the findings are shown in Figure 3. It is obvious that the initial water content significantly influenced the percentage swell.

Effect of initial water content on expansion potential of soils
Figure 3: Effect of initial water content on volume change

Chen (1973 and 1988) also performed oedometer tests on samples with different dry densities and the same initial water content. Figure 4 shows these findings for percentage swelling and swelling pressure. They demonstrate quite clearly that both the percentage swelling and the swelling pressure were significantly influenced by the initial density. As a result, the expansion potential of the soil increases with initial soil moisture content and soil density.

Effect of dry density on expansion potential of soils
Figure 4: Effect of dry density on volume change and swelling pressure

Effects of Dry Density on Expansion Potential

The expansion potential for a remoulded soil sample increases with the density to which the sample is compacted. Similar to this, the swell potential decreases as water content increases.

Dry density is one of these parameters that is crucial. This is explained by taking into account how many soil particles there are in a given volume. The surface area accessible for the adsorption of water will increase with the number of particles in a unit volume, increasing the potential for swelling.

It was observed during laboratory studies that when the degree of compaction increases, swelling in the presence of water at a certain confining pressure increases. This can be attributed to the fact that more clay particles would occupy the same volume due to the greater compaction.

Studies with compacted clays have shown that soils compacted on the dry side of the optimum moisture content are likely to have a flocculated structure, whereas soils compacted on the wet side of optimum moisture content tend to have a dispersed structure, suggesting that expansion potential may be attributable to soil structure. Although soil structure may contribute to the mechanisms causing swelling, it’s possible that the initial moisture content at which the soil is compacted is more significant.

The capacity of the soil to absorb water is closely correlated with swelling potentials. It is clear that swell potential depends on the initial moisture content because the mass can only absorb a given amount of water. Compared to samples of the same density compacted wet of optimum, samples compacted dry of optimum exhibit higher swelling characteristics and swell to larger water contents.

Who is a Geotechnical Engineer?

Geotechnical engineering is a branch of civil engineering concerned with the engineering behaviour of soils. Geotechnical engineers apply the principles of soil and rock mechanics to the design of the foundation of all structures and infrastructures resting on the ground, and the use of geomaterials (soils and rocks) for construction. They are also interested in ground improvement, and protection of the soil and groundwater from pollution. Geotechnical engineering is also known as soil engineering.

Geotechnical engineers, therefore, design the foundations of buildings, bridges, dams, embankments, highways, earthworks and deep excavations,  natural and man-made earth slopes, landfills, retaining walls, tunnels, cofferdams, etc. In practice, they work closely with structural engineers, highway/transportation engineers, geologists, water resources engineers, etc.

geotechnical engineering

Soil is a product of nature with so many different properties. Some soils are firm and stiff enough for construction purposes, while some soils are weak and marginal, hence unsuitable for construction purposes.  

The loads from all structures are ultimately transferred to the ground, and it is expected that the ground should be able to withstand the load without undergoing excessive settlement or shear failure. Furthermore, when geomaterials are to be used in civil engineering constructions such as highways, embankments, earth dams, etc, the material selected should have specific engineering properties that will prevent premature failure of the construction. It is the job of geotechnical engineers to recommend suitable materials or design the stabilisation/modification of existing materials.

During the design of foundations, the size and type of foundation can be altered to ensure that the pressure on the soil is within allowable limits. This is essentially the job of geotechnical engineers.

geotechnical engineering construction
Geotechnical engineering construction

In some cases, the load coming from the structure may be too high, or the soil near the surface may be too weak, such that the geotechnical engineer may recommend a deep foundation such as piles. Using the geotechnical soil investigation report, geotechnical engineers recommend the depth, size, type, and allowable load on the pile. 

In essence, geotechnical engineers perform geotechnical analysis to assess site condition, perform field and environmental investigations, plan and conduct geotechncial exploration, review construction design proposls and approve geotechnical aspects.

A lot of theories are applied during the design of foundations and other earthen structures, and a geotechnical engineer combines them with soil investigation findings to decide on the most suitable foundation. 

Some important topics of geotechnical engineering are;

(1) Shear strength of soils
(2) Compressibility and consolidation
(3) Stresses in soil
(4) Phase relationship and physical properties of soils 
(5) Compaction
(6) Characterisation/Classification of soils

These topics form the basis of many geotechnical engineering design concepts such as foundation design, retaining walls design, stability of slopes, dam engineering, highway construction etc. Other important topics in geotechnical engineering are the flow of water through soils (hydraulic conductivity), soil dynamics, ground improvement, etc.

Geotechnical engineers are also interested in the improvement of marginal soils such as expansive soils, collapsible soils, etc in order to improve their engineering properties. 

Depending on the findings of the geotechnical soil investigation report, overexcavation and replacement of the weak soil with a more suitable material may be recommended. Other solutions can include, stabilisation of the soil with cement, lime, pozzolans, agricultural and industrial wastes, etc. Fibres and geotextiles are also used in soil stabilisation and improvement.

Other methods such as the use of vertical sand drains, deep soil mixing, stone columns, dynamic compaction, micro piles, and grouting can be used to densify weak soils and reduce settlement potential. 

Therefore, the practice of geotechnical engineering relies heavily on soil investigations which usually include laboratory and field testing of soil and rock samples. The results obtained from laboratory investigations are then used to carry out design with an appropriate factor of safety.

Geotechnical engineers are found in private consultancy, in academia, as borehole experts, construction industry, the oil and gas industry, the mining industry, military and warfare, water resources companies, etc. Their services can include; foundation design, design of earthen structures, groundwater extraction consultancy, verification of designs, design of machine foundations, structural design of highway pavements and embankments, ground improvement, etc.

Geotechnical engineers usually have civil engineering as their first degree, and may go ahead to obtain post graduate degrees in geotechnical engineering. They must also be registered/licensed to practice engineering in their country or state. This involves passing all the professional exams and interviews involved in the qualification process.

Some of the aspects of geotechnical engineering that are distinguished as a result of research and practice are as follows:

  • Transportation geotechnics
  • Geo-environmental engineering 
  • Energy geotechnics
  • Rock mechanics
  • Soil mechanics
  • Foundation Engineering 
  • Marine/offshore geotechnics

Professional Bodies in Geotechnical Engineering

Some notable geotechnical engineering professional bodies are;